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Full text of "SOILS AND GEOLOGY PROCEDURES FOR FOUNDATION DESIGN OF BUILDINGS AND OTHER STRUCTURES (EXCEPT HYDRAULIC STRUCTURES)"

ARMY TM 5-818-1 



AIR FORCE AFM 88-3, CHAP. 7 



SOILS AND GEOLOGY 

PROCEDURES FOR FOUNDATION DESIGN OF 

BUILDINGS AND OTHER STRUCTURES 

(EXCEPT HYDRAULIC STRUCTURES) 



DEPARTMENTS OF THE ARMY AND THE AIR FORCE 

OCTOBER 1983 



This manual has been prepared by or for the Government and, except to the extent indicated below, 
is public property and not subject to copyright. 

Copyrighted material included in this manual has been used with the knowledge and permission of 
the proprietors and is acknowledged as such at point of use. Anyone wishing to make further use of any 
copyrighted material, by itself and apart from this text, should seek necessary permission directly from the 
proprietors. 

Reprint or republication of this manual should include a credit substantially as follows: "Joint 
Departments of the Army and Air Force, USA, Technical Manual TM 5-81 8-1 /AFM 88-3, Chapter 7, Soils 
and Geology Procedures for Foundation Design of Buildings and Other Structures (Except Hydraulic 
Structures), 21 October 1983." 

If the reprint or republication includes copyrighted material, the credit should also state: "Anyone 
wishing to make further use of copyrighted material, by itself and apart from this text, should seek 
necessary permission directly from the proprietors." 



TECHNICAL MANUAL 
No. 5-818-1 
AIR FORCE MANUAL 
No. 88-3, Chapter 7 



ICHAPTERI [TJ 



2. 



3. 



4- 



5. 



6. 



1-1 



1-2 



1-3 



;L4 



.LSd 



Compaction characteristics of soils I.3-2 

Density of cohesionless soils J.3-3 



Permeability 1.3-4 

Consolidation 1.3-5 

Swelling, shrinkage, and collapsibility 1 .3-6 

Shear strength of soils J. 3-7 

Elastic properties (E, u) [.3-8 



Modulus of subgrade reaction J .3-9 

Coefficient of at-rest earth pressure 1.3-10 

Properties of intact rock .13-11 



Properties of typical shales J.3-12 

FIELD EXPLORATIONS 



Investigational programs J. 4-1 



Soil boring program 1.4-2 

Field measurements of relative density and consistency I.4-3 

Boring logs L4-4 



Groundwater observations 1. 4-5 



In situ load tests L.4-6 



Geophysical exploration 1. 4-7 



Borehole surveying L.4-8 

SETTLEMENT ANALYSES 

Settlement problems 1.5-11 

Loads causing settlement 1.5-21 

Stress computations L.5-31 

Settlement of foundations on clay 1.5-41 

Consolidation settlement 1.5-51 

Settlement of cohesionless soils 1.5-61 

Eliminating, reducing, or coping with settlement 1.5-71 

BEARING-CAPACITY ANALYSIS 



W2 



Bearing capacity of soils J. 6-1 

Shear strength parameters 

Methods of analysis 

Tension forces 



Bearing capacity of rock J. 6-5 



6-3 



6-4 



This manual supersedes TM 5-81 8-1 /AFM 88-3, Chapter 7, 15 August 1961. 



*TM 5-818-1 
* AFM 88-3. Chap. 7 

HEADQUARTERS 
DEPARTMENTS OF THE ARMY 
AND THE AIR FORCE 
WASHINGTON, DC, 21 October 1983 
SOILS AND GEOLOGY 
PROCEDURES FOR FOUNDATION DESIGN OF 
BUILDINGS AND OTHER STRUCTURES 
(EXCEPT HYDRAULIC STRUCTURES) 

Paragraph Page 
INTRODUCTION 

Purpose 

Scope 

Objectives of foundation investigations 

Report of subsurface and design investigations 

IDENTIFICATION AND CLASSIFICATION OF SOIL 
AND ROCK 

Natural soil deposits 1.2-11 

Identification of soils J. 2-2 1 

Index properties L. 2-31 

Soil classification L.2-41 

Rock classification L.2-51 

Rock properties for foundation design 1.2-61 

Shales fFTl 

ENGINEERING PROPERTIES OF SOIL AND ROCK 
Scope 



1-1 



1-1 



1-1 



1-1 



23 



23] 



23 



m 



23 



Ml 



FT3l 



LM] 



1F5\ 



33 



33 



33 



3-15 



3-16 



3-22 



3-22 



3-22 



3-22 



3-22 



4-1 
4-2 
4-5 
4-6 
4-6 
4-9 
4-9 




4-121 



[S3 
E3 
LS3 

[53 

E3 
[133 



6-1 



6-1 



6-1 



6-5 



6-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



I CHAPTERI 



7. 
8. 



9. 



QH 



\12\ 

rrm 



QE 



rrm 



DEWATERING AND GROUNDWATER CONTROL 
General 



Paragraph Page 



Foundation problems 

SLOPE STABILITY ANALYSIS 
General 



Q3 

CZ2 

CS3 

Slope stability problems I.8-2I 

Slopes in soils presenting special problems I.8-3I 

Slope stability charts I.8-4I 

Detailed analyses of slope stability I.8-5I 

Stabilization of slopes I.8-6I 

SELECTION OF FOUNDATION TYPE 

Foundation-selection considerations .1.9-1 

Adverse subsurface conditions 1 .9-2 

Cost estimates and final selection 1.9-3 

SPREAD FOOTINGS AND MAT FOUNDATIONS 

General 



10-1 



10-2 



10-3 



10- 



Adequate foundation depth 

Footing design 

Mat foundations 

Special requirements for mat foundations 1. 1 0-5 

Modulus of subgrade reaction for footings and mats J. 1 0-6 

Foundations for radar towers 1.10-7 

DEEP FOUNDATIONS INCLUDING DRILLED PIERS 

General 



Floating foundations 1.11-2 



Settlements of compensated foundations J. 1 1 -3 



Underpinning 1.11- 



Excavation protection .111-5 



Drilled piers fjl 

PILE FOUNDATIONS 
General 



Design 

FOUNDATIONS ON EXPANSIVE SOILS 
General 



nsm 

r^2i 

urni 

Foundations problems 1.13-21 

RETAINING WALLS AND EXCAVATION SUPPORT 
SYSTEMS 

Design considerations for retaining walls 1.14-1 

Earth pressures 

Equivalent fluid pressures 

Design procedures for retaining walls 

Crib wall 



14-6 



14-7 



14-8 



Excavation support systems 

Struttedexcavations 

Stability of bottom of excavation 

Anchored walls 

FOUNDATIONS ON FILL AND BACKFILLING 

Types of fill QFj 



14-2 



TT3" 



14-4 



14-5 



14-9 



Foundations on compacted fills 1 .15-2 

Compaction requirements .15-3 

Placing and control of backfill J. 15-4 



Fill settlements L. 1 5-5 

Hydraulic fills UE 

STABILIZATION OF SUBGRADE SOILS 

General 1.16-11 

Vibrocompaction L. 16-21 

Vibrodisplacement compaction 1.16-31 

Grouting and injection J. 1 6-4I 

Precompression 1.16-51 

Reinforcement 1.16-61 

Miscellaneous methods 1.16-71 

DESIGN FOR EQUIPMENT VIBRATIONS AND SEISMIC 
LOADINGS 

Introduction 1.17-11 

Single degree of freedom, damped, forced systems 1.17-21 

Foundations on elastic soils 1.17-31 



73 

13 



83 

13 
E3 

W3\ 

ED 



9-1 
9-2 



10-1 


10-1 


10-1 


10-1 


10-5 


10-5 


10-5 



11 


-1 


11 


-1 


11 


-1 


11 


-3 


11 


-5 


11 


-9 



Q23 
□23 

□33 
Q33 



14-1 




14-1 




14-1 




14-5 




14-9 




14-9 




14-11 




14-17 


14-19 



15-1 



15-1 



15-2 



15-2 



15^41 



15-41 



QE3] 
□13 
rr^roi 

116-121 
116-141 
116-161 
116-161 



Q73 
QZ3 

□ZEU 



TM 5-818-1 / AFM 88-3, Chap. 7 



I CHAPTERI HE 



I APPbNDIXI Qg 



Figure 



2-1 1 

2-2. 

2-3. 

2-4. 

2-5. 

3-1 

3-2 

3-3 
3-4 

I 3-5I 

3-6. 

3-7. 

I3-8I 

3-9I 

3-11 
3-1 ' 

3-1 £ 

3-1 ; 

3-1' 

3-1 J 

3-K 
3-1" 

3-1! 

3-K 
3-2( 


)\ 
I 

I. 

3. 

i 

1 

3 

3, 
). 


3-2' 

4-1, 
4-2 

4-3. 
4-4. 
4-5. 

4-6 

5-1 
5-2 

5-3 
5-4. 
5-5. 


! 



Pa ragrap h 

Wave transmission, attenuation, and isolation 1. 17-41 

Evaluation of S-wave velocity in soils 1.17-51 

Settlement and liquefaction J. 17-61 

Seismic effects on foundations 1.17-71 

FOUNDATIONS IN AREAS OF SIGNIFICANT FROST 
PENETRATION 

Introduction 1.18-11 

Factors affecting design of foundations 1.18-21 

Site investigations 1.18-31 

Foundation design L. 18-41 

REFERENCES , , 

I LIST OF FIGURESl 



Page 
QZal 

rrrrai 



DEB 
rTF5i 
rresi 
EJJEB 



Page 



2-4 



2-6 



2-8 



2-15 



2-16 
T5[ 



Distribution of natural soil deposits in the United States 

Typical grain-size distribution curves 

Weight-volume relationships 

Modified core recovery as in index of rock quality 

Classification of shales 

Typical CE 55 compaction test data 

Relation between relative density and dry density (scaled to plot 

as a straight line) |. 3-4 

Angle of friction versus dry density for coarse-grained soils . [ 3-4 

Generalized curves for estimating emax and em,, from gradational 

and particle shape characteristics I. 3-5I 

A summary of soil permeabilities and method of determination L 3-6I 

Examples of laboratory consolidation test data . 1 3-71 

Analyses of consolidation test data I. 3-81 

Approximate relation between liquidity index and effective overburden 

pressure, as a function of the sensitivity of the soil 13-91 

Approximate relation between void ratio and effective overburden pres- 
sure for clay sediments, as a function of the Atterberg limits I 3-101 

Approximate correlations for swelling index of silts and clays L 3-131 

Correlations between coefficient of consolidation and liquid limit L 3-141 

Predicted relationship between swelling potential and plasticity index 

for compacted soils I 3-151 

Guide to collapsibility, compressibility, and expansion based on in situ 

dry density and liquid limit I 3-161 

Shear test apparatus and shearing resistance of soils I 3-171 

Remolded shear strength versus liquidity index relationships for differ- 



3-23 



3-24 



entclays 173-18 

Normalized variation of su/po ratio for overconsolidated clay . I 3-19 

General relationship between sensitivity, liquidity index, and effective 

overburden pressure I. 3-211 

Empirical correlation between friction angle and plasticity index from 

triaxial compression tests on normally consolidated undisturbed I 3-221 

clays 

Relation between residual friction angle and plasticity index 

Chart for estimating undrained modulus of clay 

Coefficient of earth pressure at rest (Ko) as a function of overconsolidat- 
ed ratio and plasticity index | 3-251 

Relative density of sand from the standard penetration test . [ 4-51 

Rough correlation between effective friction angle, standard blow, 

count and effective overburden pressure 1 4-61 

Correction factor for vane strength I 4-71 

Typical log of boring L 4-81 

Typical details of Casagrande piezometer and piezometer using well 

screen I 4-91 

Determination of permeability from field pumping test on a fully 

pene-trating well in an artesian aquifer I 4-101 

Vertical stress beneath a uniformly loaded rectangular area J. 5-51 

Vertical stress beneath a triangular distribution of load on a rectangular 

area I 5-61 

Influence value for vertical stress under a uniformly loaded circular area I. 5-71 

Example of settlement analysis I 5-81 

Time factors for various boundary conditions 1 5-91 

Ultimate bearing capacity of shallow foundations under vertical, eccen- 
tric loads 1 6-31 

iii 



TM 5-818-1 / AFM 88-3, Chap. 7 



Figure 



E3 



53] 



ES 



Mi 



83 



F21 



E31 



8-41 



8-5. 


8-6. 


8-7 


8-8.| 


8-9 


8-10 


8-11 



8^T?1 



TM] 

TEa] 
nz] 



133] 



TT31 



T4TI 
TT21 
TT31 
14^4] 



14-51 


114-61 
TT9T 


14-10] 


14-111 


14-12] 


14-13.1 


14-14] 


16-11 


16-a. 




16-31 


16-41 


16-51 




16-6.1 




17-11 


17-21 




17-31 


17-41 




17-51 


17-61 


17-71 



Ultimate bearing capacity with groundwater effect 

Ultimate bearing capacity of shallow foundations under eccentric inclined 
loads 



Page 
..CE3I 

...ESI 



Example of bearing capacity computation for inclined eccentric load on 

rectangular footing L 6-61 

Ultimate bearing capacity of deep foundations L 6-71 

Bearing capacity factors for strip and circular footings on layered founda- 

tions in clay ..I 6-81 

Slope stability charts for = soils J. 8-41 

Reduction factors ( |i q , u w and _'w) for slope stability charts for = 

and > soils I. 8-51 

Reduction factors (tension cracks, It) for slope stability charts for =0 

and > soils I. 8-61 

Examples of use of charts for slopes in soils with uniform strength and 

= 



8-8 



8-9 



8-10 



8-11 



Slope stability charts for + > soils 

Stability charts for infinite slopes 

Slope stability charts for t = and strength increasing with depth 

Method of moments for = 

Example problem for ordinary method of slices 

Example of use of tabular form for computing weights of slices 

Example of use of tabular form for calculating factor of safety by ordi- 
nary method of slices J~8-14 

Example of simplified wedge analysis I. 8-15 



8-7 



8-12 



8-13 



10-11 Example of method for selecting allowable bearing pressure .L 10-2 



Proportioning footings on cohesionless soils 1 1 0-3 



TT5" 



14-6 



VF7 



14-8 



14-9 



14-10 



14-11 



14-17 



14-18 



14-19 



Distribution of bearing pressures 1. 10-4 

Effect of pore pressure dissipation during excavation and settlement re- 
sponse 1 1 1 -21 

Excavation rebound versus excavation depth 11 1-31 

Probable settlements adjacent to open cuts 111 -41 

Methods of underpinning .1 1 1 -51 

Load diagrams for retaining walls I. 14-21 

Active pressure of sand with planar boundaries L 14-31 

Passive pressure of sand with planar boundaries 1 14-41 

Active and passive earth pressure coefficients according to Coulomb 

theory 

Horizontal pressures on walls due to surcharge 

Design loads for low retaining walls, straight slope backfil 

Design loads for low retaining walls, broken slope backfill 

Estimates of increased pressure induced by compaction 

Design criteria for crib and bin walls 

Types of support systems for excavation 

Pressure distribution-complete excavation 

Stability of bottom of excavation in clay 

Typical tieback details 

Methods of calculating anchor capacities in soil f 1 4-20 

Applicable grain-size ranges for different stabilization methods . [ 1 6-8I 

Range of particle-size distributions suitable for densification by vibro- 

compaction ] 16-91 

Sand densification using vibratory rollers . 116-111 

Relative density as a function of vibroflot hole spaci ngs 1 16-121 

Allowable bearing pressure on cohesionless soil layers stabilized by vibro- 

flotation 116-131 

Soil particle sizes suitable for different grout types and several concen- 
trations and viscosities shown 116-1 51 

Response spectra for vibration limits 1. 1 7-2I 

Response curves for the single-degree-of-freedom system with viscous 

damping L 17-31 

Six modes of vibration for a foundation 1 17-31 

Equivalent damping ratio for oscillation of rigid circular footing on elas- 
tic half-space 1. 17-51 

Examples of computations for vertical and rocking motions L 1 7-6I 

Coupled rocking and sliding motion of foundation 1 17-81 

Distribution of displacement waves from a circular footing on the elastic 

half-space 1. 17-101 



IV 



TM 5-818-1 / AFM 88-3, Chap. 7 



Figure 



Table 



QZS] 

rrm 
QZ3S 



□seu 

HTF31 
HTF51 



2-2. 
2-3. 
2-4 
2-5. 
2-6. 
2-7. 

33] 

[33] 

[33] 
[33] 

[33] 

E3] 
B2I 
E3I 

337 

5-1. 

M 

mi 

5^4] 
6-1 

331 
331 
[33] 

TN2l 
T4TI 
1T21 
1T31 

3Q 

1F21 

3B31 

TS^2l 
Q13I 



Page 

Theoretical relation between shear velocity ratio VpN,V and Poisson's .117-111 

Idealized cyclic stress-strain loop for soil 1 17-121 

Variation of shear modulus with cyclic strain amplitude; G.x = G at 

= 1 to 3 x 10-' percent; scatter in data up to about +0.1 on vertical scale 117-151 

Variation of viscous damping with cyclic strain amplitude; data scat - 

ter up to about ± 50 percent of average damping values shown for any 

strain 17-161 

Frost and permafrost in North America 1 18-21 

Ground temperatures during freezing season in Limestone, Maine 1 18-41 

Ground temperatures during freezing season in Fairbanks, Alaska 1 1 8-4I 

Design alternatives J. 18-71 

Heave force tests, average tangential adfreeze bond stress versus time , 
and timber and steel pipe piles placed with silt-water slurry in dry 
excavated holes. Piles were installed within annual frost zone only, over 

permafrost, to depths from ground surface of 3.6 to 6.5 feet 118-111 

pJST OF TABLES] 

Page 



2-2 



2-5 



2-7 



2-9 



2-10 



2-11 



14 

33T 



333] 



333 



A Simplified Classification of Natural Soil Deposits 

Determination of the Consistency of Clays 

Unified Soil Classification System 

Descriptive Soil Names Used in Local Areas (L) and Names Widely Used 

A Simplified Classification for Rocks 

Descriptive Criteria for Rock 

Engineering Classification of Intact Rock 

Typical Engineering Properties of Compacted Materials 

Estimating Degree of Preconsolidation 

Compression Index Correlations 

Value of Coefficient of Compressibility (mv) for Several Granular Soils 

During Virgin Loading I. 3-121 

Sensitivity of Clays I. 3-201 

Values of Modulus of Subgrade Reaction (ks) for Footings as a Guide 

to Order of Magnitude I. 3-241 

An Engineering Evaluation of Shales I. 3-261 

Physical Properties of Various Shales I. 3-271 

Types and Sources of Documentary Evidence .[ 4-1 1 

Recommended Undisturbed Sample Diameters J. 4-3I 

Recommended Minimum Quantity of Material for General Sample Lab- 

oratory Testing I. 4-4I 

Surface Geophysical Methods I. 4-111 

Borehole Surveying Devices 1.4-131 

Causes of Settlement I. 5-21 

Values of Angular Distortion (5/i) That Can Be Tolerated Without 

Cracking J. 5-31 

Empirical Correlations Between Maximum (A) and Angular Distortion 



(5/i). 



.IT^ 



Methods of Eliminating, Reducing, or Coping with Settlements I 5-111 

Estimates of Allowable Bearing Pressur |. 6-2 1 

Allowable Bearing Pressure for Jointed Rock . ] 6-9J 



Methods of Stabilizing Slopes and Landslides J. 8-161 

Foundation Possibilities for Different Subsoil Conditions I 9-11 

Checklist for Influence of Site Characteristics on Foundation Selection 

for Family Housing I. 9-31 

Excavation Protection [ 11-61 

Design Parameters for Drilled Piers in Clay . 1 11-101 

Types of Walls J 14-121 

Factors Involved in Choice of a Support System for a Deep Excavation 1 14-141 

Design Considerations for Braced and Tieback Walls L 14-151 

A Summary of Densification Methods for Building Foundations L 1 5-3I 

Compaction Density as a Percent of CE 55 Laboratory Test Density 1 1 5-4I 

Stabilization of Soils for Foundations of Structures L 16-21 

Applicability of Foundation Soil Improvement for Different Structures 

and Soil Types (for Efficient Use of Shallow Foundations) 1 16-71 

Vibroflotation Patterns for Isolated Footings fo r an Allowable Bearing 

Pressure J. 16-131 



TM 5-818-1 / AFM 883, Chap. 7 



Table 



Page 
1 17-1.1 Mass Ratio, Damping Ratio, and Spring Constant for Rigid Circular Foot- 
ing on the Semi-Infinite Elastic Body [17-4 1 

TTzl Values of kL/k for Elastic Layer (k from Table 17-1) mTTI 

17-3. Attenuation Coefficients for Earth Materials 117-101 

17-4.1 Values of Constant rYUsed with Equation (17-23) to Estimate Cyclic 

Shear Modulus at Low Strains 11 7-131 

1 17-5.1 Values of Constant K, Used with Equation (17-24) to Estimate Cyclic 

Shear Modulus at Low Strains for Sands [17-141 

1 17-6.1 Criteria for Excluding Need for Detailed Liquefaction Analyses J 17-1 71 



VI 



TM 541-1 1 / AFM 883. Chap. 7 



CHAPTER 1 



INTRODUCTION 



1-1. Purpose. This manual presents guidance for 
selecting and designing foundations and associated 
features for buildings, retaining structures, and 
machinery. Foundations for hydraulic structures are not 
included. Foundation design differs considerably from 
design of other elements of a structure because of the 
interaction between the structure and the supporting 
medium (soil and/or rock). 

1-2. Scope. Information contained herein is directed 
toward construction usually undertaken on military 
reservations, although it is sufficiently general to permit 
its use on a wide variety of construction projects. 

a. This manual includes- 

(1) A brief summary of fundamental 
volumetric - gravimetric relationships. 

(2) Summaries of physical and engineering 
properties of soil and rock. 

(3) General descriptions of field and 
laboratory investigations useful for foundation selection 
and design. 

(4) Design procedures for construction 
aspects, such as excavated slopes and shoring. 

(5) Empirical design equations and 
simplified methods of analysis, including design charts, 
soil property-index correlations, and tabulated data. 

(6) Selected design examples to illustrate 
use of the analytical methods. 

b. Since the user is assumed to have some 
familiarity with geotechnical engineering, design equa- 
tions and procedures are presented with a minimum of 
theoretical background and no derivations. The topics of 
dewatering and groundwater control, pile foundations, 
and foundations on expansive soils are covered in 
greater depth in separate technical manuals and are only 
treated briefly in this manual. 

1-3. Objectives of foundation investigations. The 

objectives of foundation investigations are to determine 
the stratigraphy and nature of subsurface materials and 
their expected behavior under structure loadings and to 
permit savings in design and construction costs. The 
investigation is expected to re- veal adverse subsurface 
conditions that could lead to construction difficulties, 
excessive maintenance, or possible failure of the 
structure. The scope of investigations depends on the 
nature and complexity of sub-surface materials and the 
size, requirements for, and cost of the structure. 



1-4. Report of subsurface and design 
investigations. The report should contain sufficient 
description of field and laboratory investigation, sub- 
surface conditions, typical test data, basic assumptions, 
and analytical procedures to permit detailed re- view of 
the conclusions, recommendations, and final design. The 
amount and type of information to be presented shall be 
consistent with the scope of the investigation. For some 
structures, a cursory review of foundation conditions may 
be adequate. For major structures, the following outline 
shall be used as a guide: 

a. A general description of the site, indicating 
principal topographic features in the vicinity. A plan map 
that shows the surface contours, the location of the 
proposed structure, and the location of all borings should 
be included. 

b. A description of the general and local 
geology of the site, including the results of the geological 
studies. 

c. The results of field investigations, including 
graphic logs of all foundation and borrow borings, 
locations of and pertinent data from piezometers, and a 
general description of subsurface materials, based on the 
borings. The information shall be presented in 
accordance with Government standards. The boring logs 
should indicate how the borings were made, type of 
sampler used, split-spoon penetration resistance, and 
other field measurement data. 

d. Groundwater conditions, including data on 
seasonal variations in groundwater level and results of 
field pumping tests, if performed. 

e. A general description of laboratory tests 
performed, range of test values, and detailed test data on 
representative samples. Atterberg limits should be 
plotted on a plasticity chart, and typical grain-size curves 
on a grain-size distribution plot. Laboratory test data 
should be summarized in tables and figures as 
appropriate. If laboratory tests were not performed, the 
basis for determining soil or rock properties should be 
presented, such as correlations or reference to pertinent 
publications. 

f. A generalized geologic profile used for 
design, showing properties of subsurface materials and 
design values of shear strength for each critical stratum. 
The profile may be described or shown graphically. 



1-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



g. Where alternative foundation designs are 
prepared, types of foundations considered, together with 
evaluation and cost data for each. 

h. A table or sketch showing the final size and 
depth of footings or mats and lengths and types of piles, 
if used. 

/. Basic assumptions for loadings and the 
computed factors of safety fo r bearing-capacity 
calculations, as outlined i n| chapter"6l 

j. Basic assumptions, loadings, and results of 
settlement analyses, as outlined in chapters 5, 6, and 10; 



also, estimated swelling of subgrade soils. The effects of 
computed differential settlements, and also the ef- fects 
of swell, on the structure should be discussed. 

k. Basic assumptions and results of other 
analyses. 

/. An estimate of dewatering requirements, if 
necessary. The maximum anticipated pumping rate and 
flow per foot of drawdown should be presented. 

m. Special precautions and recommendations 
for construction. Possible sources for fill and backfill 
should also be given. Compaction requirements should 
be described. 



1-2 



TM 5-81 8-1 /AFM 88-3. Chap. 7 



CHAPTER 2 



IDENTIFICATION AND CLASSIFICATION OF SOIL AND ROCK 



2-1. Natural soil deposits. 

a. The character of natural soil deposits is 
influenced primarily by parent material and climate. The 
parent material is generally rock but may include partially 
indurated materials intermediate between soil and rock. 
Soils are the results of weathering, mechanical 
disintegration, and chemical decomposition of the parent 
material. The products of weathering may have the 
same composition as the parent material, or they may be 
new minerals that have resulted from the action of water, 
carbon dioxide, and organic acids with minerals 
comprising the parent material. 

b. The products of weathering that remain in 
place are termed residual soils. In relatively flat regions, 
large and deep deposits of residual soils may 
accumulate; however, in most cases gravity and erosion 
by ice, wind, and water move these soils to form new 
deposits, termed transported soils. During 
transportation, weathered material may be mixed with 
others of different origin. They may be ground up or 
decomposed still further and are usually sorted according 
to grain size before finally being deposited. The newly 
formed soil deposit may be again subject to weathering, 
especially when the soil particles find themselves in a 
completely different environment from that in which they 
were formed. In humid and tropical climates, weathering 
may significantly affect the character of the soil to great 
depths, while in temperate climates it produces a soil 
profile that primarily affects the character of surface soils. 

c. The character of natural soil deposits 
usually is complex. A simplified classification of natural 
soil deposit s based on methods of deposition is given in 

I table 2^il , together with pertinent engineering 
characteristics of each type. More complete descriptions 
of natural soil deposits are given in geology textbooks. 
The highly generalized map in I figure 2-1 I shows the 
distribution of the more important natural soil deposits in 
the United States. 

2-2. Identification of soils. 

a. It is essential to identify accurately materials 
comprising foundation strata. Soils are identified by 
visual examination and by means of their index 
properties (grain-size distribution, Atterberg limits, water 
content, specific gravity, and void ratio). A description 
based on visual examination should include color, odor 
when present, size and shape of grains, gradation, and 
density and consistency characteristics.- Coarsegrained 



soils have more than 50 percent by weight retained on 
the No. 200 sieve and are described primarily on the 
basis of grain size and density. With regard to grain-size 
distribution, these soils should be described as uniform, 
or well-graded; and, if in their natural state, as loose, 
medium, or dense. The shape of the grains and the 
presence of foreign materials, such as mica or organic 
matter, should be noted. 

b. Fine-grained soils have more than 50 
percent by weight finer than the No. 200 sieve. 
Descriptions of these soils should state the color, texture, 
stratification, and odor, and whether the soils are soft, 
firm, or stiff, intact or fissured. The visual examination 
should be accompanied by estimated or laboratory- 
determined index properties. A summary of expedient 
tests for identifying fine-grained soils is given in ltable 2-21 
The important index properties are summarized in the 
following paragraphs. Laboratory tests for determining 
index properties should be made in accordance with 
standard procedures. 

2-3. Index properties. 

a. Grain-size distribution. The grain-size 
distribution of soils is determined by means of sieves 
and/or a hydrometer analysis, and the results are 
expressed in the form of a cumulative semilog plot of 
percentage finer versus grain diamete r. Typical g rain- 
size distribution curves are shown in figure 2-2~| The 
knowledge of particle-size distribution is of particular 
importance when coarse-grained soils are involved. 
Useful values are the effective size, which is defined as 
the grain diameter corresponding to the 10 percent finer 
ordinate on the grain-size curve; the coefficient of 
uniformity, w hich is d efined as the ratio of the D6oSize to 
the D, , size Kfig 2-2)1 : the coefficient of curvature, which 
is defined as the ratio of the s quare of th e D0 3 size to the 
product of the D i a n d n Rn sizes Ifiable 2-3|) : and the 15 and 
85 percent sizes, which are used in filter design. 

b. Atterberg limits. The Atterberg limits 
indicate the range of water content over which a cohesive 
soil behaves plastically. The upper limit of this range is 
known as the liquid limit (LL); the lower, as the plastic 
limit (PL). The LL is the water content at which a soil will 
just begin to flow when slightly jarred in a pre 



2-1 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 2-1. A Simplified Classification of Natural Soil Deposits 



Major Division* 



Principal Soil Type 



Material formed by 
disintegration of 
underlying parent 
rock or partial ly 
indurated Material! 



Accumulations of 

highly organic Mate- 
rial formed in pLace 
by the growth and 
auhsequent decay of 
plant life 

Mater ia 1 transported 
and deposited by 
running water 



Material deposited 
in a lake 



Material deposited 
an estuary 



Residual sands and rock fragment a of various size* forawd 
by solution and leaching of cementing arterial, leaving 
the store resistant particles; coaasonly quartz 



Pertinent Engineering Characteristics 

Generally favorable foundation 
conditions . 



Residual clay , extremely finely divided clay Material fonsed Variable properties requiring mvesti- 

in place by the weathering of rock, derived either by the gat ion to determine depth and conditon 

chemical decay of feldspar and other rock Minerals or by of weathering. 
the removal of nonclay-Mineral constituents by aolution 
froM a clay-bearing rock. 



Peat . A somewhat fibrous aggregate of decayed and decaying 
vegetable Matter having a dark color and odor of decay 

Muck. Finely divided, wel 1 -decomposed organic Material 
intermixed with a high percentage (20-50%) of Mineral 
■atter 

Floodplain deposits - Unconsolidated soils deposited by a 
stream within that portion of its valley subject to in-, 
undation by floodwater 

Natural levees. Long, broad, low ridges of sand, silt, or 
silty clay deposited by a stream on its floodplain and 
along both banks of its channel during overbank flow. 

Point bar . Alternating deposits of arcuate ridges and 
swales (lows) formed on the inside or convex bank of 
Migrating river bends. Ridge deposits consist prisurily 
of silt and sand, swales are clay-filled 



Channel fill . Deposits laid down in abandoned meander loops 
isolated when rivers shorten their courses. Composed pri- 
marily of clay; however, silty and ssndy soils are found 
at the upstream and downstream ends 

Backawamp . The prolonged accumulation of floodwater sedi- 
ments in flood basins behind the natural levees of a 
river. Materials are generally clays but tend to become 
More silty near riverbank 



Very compressible. Entirely unsuitable 
for supporting building foundations 



Generally favorable foundation 
conditions. 



Generally favorable foundation condi- 
tions; however, detailed investiga- 
tions are necessary to locate discon- 
tinuities. Flow slides may be a 
problem along riverbanks. 

Fine-grained soils are usually compress- 
ible. Portions may be very heteroge- 
neous. Silty soils generally present 
favorable foundation conditions 

Relatively uniform in a horizontal direc- 
tion. Clays sre usually subjected to 
sessooal volume changes 



Terrace deposits . Unconsolidated alluvium (including gravel) Generally favorable conditions, 
produced by renewed dovneutting of the valley flood by a not subject to flooding, 
rejuvenated stream 



Usually 



Fan Deposits . Alluvial deposits at foot of hills or Moun- 
tains. Extensive plains or alluvial fans 

Deltaic deposits . Deposits formed at the mouths of rivers 
which result in extension of the shoreline 

Lacustrine deposits . Material deposited within lakes (other 
tbsn those associated with glsciation) by waves, currents, 
and and organo-chemical processes. Deposits consist of 
unstratified organic clay or clay in central portions of 
the lake and typically grade to stratified silts and sands 
in peripheral zones 

Eatuarine deposits . Fine-grained sediment (usually silt 
and clay) of Marine and fluvial origin Mixed with decom- 
posed organic matter laid down in brackish water of an 
estuary 



Generally favorable foundation 
conditions 

Generally fine-grained and compressible. 
Many local variations in soil condition 

Usually very uniform in horizontal direc- 
tion. Fine-grained soils generally 
compressible 



Generally compressible, 
varations 



Many local 



Material transported 
and deposited by 
wind 



Material transported 
and deposited by 
glaciers, or by 
melt water from the 
glacier 



Loess . An unstratified calcerous deposit consisting pre- 
dominantly of silt with subordinate grain sizes ranging 
from sand to clay. Often contains foasils and is tran- 
versed by a network of small, narrow, vertical tubes 
frequently filled with calcium carbonate conceptions 
formed by root fibers now decayed. 

Dune sands . Mounds, ridges, and hills of uniform fine sand 
characteristically exhibiting rounded grains 

Glacial till . An accumulation of debris, deposited beneath, 
at the side (lateral moraines), or at the lower linit of 
a glacier (terminal moraine). Material lowered to ground 
surface in an irregular sheet by a melting glacier is 
known as a ground Moraine 



Glacio-Fluvial deposits . Coarse and fine-grained Material 
deposited by streams of melt water from glaciers. Mate- 
rial deposited on ground surface beyond terminal of 
glacier is known as an outwash plain. Gravel ridges 
known as kames and eskers 

Glacio-Lacuatrine deposits . Material deposited within lakes 
by melt water froM glaciers. Consisting of clay in cen- 
tral portions of lake and alternate layers of silty clay 
or silt and clay (varved clay) in peripheral zones 

(Continued) 



Relatively uniform deposits character- 
ized by ability to stand in vertical 
cuts. Collapsible structure. Deep 
weathering or saturation can modify 
characteristics 



Very iniform grain-size; auy exist in 
relatively loose condition 

Consists of material of all sizes in 
various proportions from boulders and 
grsvel to clay. Deposits are unstrati- 
fied. Generally present favorable 
foundation conditiona; but, rapid 
changws in conditions are common 

Many local variations- Generally ore- 
sent fsvorable foundation conditions 



Very uniform in s horizontal direction 



2-2 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 2-1. A Simplified Classification of Natural Soil Deposits-Continued 



Major Divisions 



Principal Soil Type 



Material transported 
and deposited by 
ocean waves and cur- 
rents in shore and 
offshore areas 



Material transported 
and deposited by 
gravity 



Material ejected from 
volcanoes and trans- 
ported by gravity 
wind, and air 



Shore deposits . Deposits of sands and/or gravels formed by 
the transporting, destructive, and sorting action of 
waves on the shoreline 

Marine clays . Organic and inorganic deposits of fin- 
grained Material 



Talus . Deposits created by gradual accumulation of un- 
sorted rock fragments and debris at base of cliffs 

Colluvial deposits . Fine colluviusj consisting of clayey 
sand, sandy silt, or clay 

LandBide deposits . Considerable susses of soil or rock 
that have slipped down, sore or less as units, fron their 
former position on steep slopes 

E.jecta . Loose deposits of volcanic ash, lapilli, boabs, 
cinders, etc 



Pertinent Engineering Characteristics 

Relatively uniform and of moderate to 
high density 



Generally very uniform in composition. 
Compressible and unually very sensi- 
tive to remolding 



Previous movements indicate possible 
future difficulties. Generally un- 
stable foundation conditions 



Typically shardlike particles of silt 
size with larger volcanic debris. 
Weathering and redeposition produce 
highly plastic compressible clay. 
Unusual and difficult foundation 
conditions 



U. S. Army Corps of Engineers 

scribed manner. The PL is the water content at which the 
soil will just begin to crumble when rolled into threads 1/8 
inch in diameter. Fat clays that have a high content of 
colloidal particles have a high LL, while lean clays having 
a low content of colloidal parti- cles have a 
correspondingly low LL. A decrease in LL and PL after 
either oven- or air-drying usually indicates presence of 
organic matter. The plasticity index (PI) is defined as 
the difference between the LL and PL. The liquidity 
index (LI) is defined as the natural water content w n , 
minus the PL, divided by the PI; i. e. , LI = (w n - PL)/PI. 
The LI is a measure of the consistency of the soils. Soft 
clays have an LI approaching 100 percent; whereas, stiff 
clays have an LI approaching zero. 

c. Activity. The activity, A, of a soil is defined 
as A = PI/(% < 0.002 mm). The activity is a useful 
parameter for correlating engineering properties of soil. 

d. Natural water content. The natural water 
content of a soil is defined as the weight of water in the 
soil expressed as a percentage of dry weight of solid 
matter present in the soil. The water content is based 
on the loss of water at an arbitrary drying temperature of 
1050to1100C. 

e. Density. The mass density of a soil material 
is its weight per unit volume. The dry density of a soil is 
defined as the weight of solids contained in the unit 
volume of the soil and is usually expressed in pounds per 
cubic foot. Various weight-volume relationships are 
presented ir l figure 2-"3l 



f. Specific gravity. The specific gravity of the 
solid constituents of a soil is the ratio of the unit weight of 
the solid constituents to the unit weight of water. For 
routine analyses, the specific gravity of sands and clayey 
soils may be taken as 2. 65 and 2. 70, respectively. 

g. Relative density. Relative density is 
defined by the following equation: 



DR(%) 



e X100 



(2-1) 



Omax " Smin 

where 

emax = void ratio of soil in its loosest state 
e = void ratio in its natural state 

emin = void ratio in its densest possible state 
Alternatively, 

D R (%) = yd-yd mi n X yd max X 1 00 (2-2) 

ydmax " ydmin "^ 

where 

yd = dry unit weight of soil in its natural 
state 
ydmin = dry unit weight of soil in its loosest 

state 
ydmax = dry unit weight of soil in its densest 
state 
Thus, DR = 100 for a very dense soil, and DR = for a 
very loose soil. Methods for determining e max or cor- 
responding densities have been standardized. Relative 
density is significant only in the case of coarse-grained 
soils. 



2-3 



TM 5-818-1 / AFM 88-3, Chap. 7 







ttottnun Man.* 



i — | 01*110*10 * 



IHIWWTW jQH.1 



EZ2 



C 1 ^^ 



E1E3 



Frara 



Figure 2-1. Distribution of natural soil deposits In the United States. 



2-4 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 2-2. Determination of the Consistency of Clays 



Unconfined 
Compressive 
Strength, q u 
tsf 

Less than 0.25 

0.25 - 0.5 

0.5-1.0 

1.0-2.0 

2.0-4.0 
Over 4.0 



Field Identification 



Easily penetrated several inches by fist 

Easily penetrated several inches by 
thumb 

Can be penetrated several inches by 
thumb with moderate effort 

Readily indented by thumb but pene- 
trated only with great effort 

Readily indented by thumbnail 

Indented with difficulty by thumbnail 



Consistency 
Very soft 
Soft 

Medium 

Stiff 

Very stiff 
Hard 



U. S. Army Corps of Engineers 



h. 



Consistency. The consistency of an 
undisturbed cohesive soil may be expressed 
quantitatively by the unconfined compressive strength 
qu. Qualitative expressions for the consistency of clays in 
terms of q^ are given ir l table 2^ . If equipment for 
making unconfined compression tests is not available, a 
rough estimate can be based on the simple field 
identification suggested in the table; various small 
penetration or vane devices are also helpful. 

2-4. Soil classification. The Unified Soil Classi- 
fication System, based on identification of soils ac- 
cording to their grain-size distribution, their plasticity 
characteristics, and their grouping with respect to be- 
havior, should be u sed to classi fy soils in connection with 
foundation design. I Table 2-3l summarizes the Unified 
Soil Classification System and also presents field 
identification procedures for fine-grained soils or soil 
fractions. It is generally advantageous to include with the 
soil classification anv regional or locally accepted 
terminology as well as the soil name ftable 2-41 . 
2-5. Rock classification. 

a. Geological classification. The geological 
classification of rock is complex, and for most 
engineering applications a simplified system of 



classification, as shown in I table 23] will be adequate. 
For any in-depth geology study, proper stratigraphic 
classification by a qualified geologist should be made to 
ensure that proper interpret ation of p rofiles is being 
made. All the rock types ir l table 2 : 5l may exist in a 
sound condition or may be fissured, jointed, or altered by 
weathering to an extent that will affect their engineering 
behavior. Descripti ve criteria for the field classification of 
rock is contained in l table 2-61 

b. Classification of intact rock. An enginee ring 
classification of intact rock is contained in ltable 2-71 The 
classification is based on the uniaxial compressive 
strength and the tangent modulus. 

2-6. Rock properties for foundation design. 

a. The principal rock properties of concern for 

foundation design are the structural features and shear 

strength. Strength properties of rock are discussed in 

I chapter 3l Structural features include- 

(1) Types and patterns for rock defects Ifiablel 
1 2-6) -cracks, joints, fissures, etc. 

(2) Bedding planes-stratification and slope 
(strike and dip). 

(3) Foliation-a general term for a planar 
arrange- 



2-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



fit 

I" 



/ , i , f t i ll 



U. S Standard Sana Numbart 



Hydromatar 



NO. 
1 

2 



CLASSIFICATION 



<=U 




GRAY, FIRM, SILT 

DARK GRAY, SOFT. 

FAT CLAY 

OARKGRAY, MEDIUM, CL 

LEAN CLAY 

TAN, UNIFORM SP 

FINE SAND 



EXAMPLE CALCULATION OF 
COEFFICIENT OF UNIFORMITY 
FOR SOIL NO. 4 



SRAVtL 



U. S. Army Corps of Engineers 



Figure 2-2. Typicalgrain-size distribution curves. 



2-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



si 

ii. 

!I3 



ir 

■ s 

s « 
- s 

5 - 

a! i 

ills 
I" ^ 

S3 i! 



1 



2 = i 

is : 



il 1, 



V- f. 

£ S .{ 

.J S, 

S" V 

Is* I 

'in 



5 5- 

l c |'s 



m 



i £ i« 



it 



Highly Organic Solli 



Poorly |nil«] £"" 



SUty gravels, grawl-sa 



Clayey gravels, gra™l-sand.clay alxturai. 



d sands, pwilly Hilda, llttla 



Poorly graded hMi or aravelly eandii 



Inorganic ultt and «ry fin* sands, ro 
Hour, silly or clayey flna lands or 
clayey allts with sllgat plaatlclty. 



norganlc clay* of low to aajdtw plMUclty, 
grenlly clays, sandy claya, sllty clays, 
lean clays. 



organic clays of also plasticity, fat clayi 



Organic clays of «rdi™ to nigh p 



Field Identification Procedures 
(Excluding particles larger than ) In. 
ana baaing fractions on eatlaiatcd weight t ] 



Wide range In grain ai»a and a 



lntarsisdlate alias alas 



Dry Strength El latency Toughness 

(Cniahlng (Reaction (ContlHewj 

stlea) to shaking) aw PL] 



Tor undisturbed soils add 



grain*; local ur geoluglc 



Sllty si 



gravel particles 1/S-in. 

sand grains, coarse lo fine; about 15* 
iwvplaatlc finals vlth lov dry strength; 



Par undisturb*d soil* add InforaalU 
on structure., stratification, con- 
sistency In undlaturbed and re. 
■elded states, BBlsture and drain. 



indicate degree and 



dascrlptl™ lnfor^tlon; and sy^xil 
In parentheses. 



Clayey Mlt, brovn; slightly plastic; 
samll percentage of fin* sand; 
nuBtroui rertleal root twin: Tin 
and dry In place; Ineaa, [ML)- 



lis 

Is- 


i 
I 


ii\ 


Sfi s 


"1 

53S 


&S1 


:;i 


*B 


III 


»! 



<v 



Mot —atlng all gradation 



Atterberg limit. 



- Greater than i, 



<y_ 






lot BMtlng all gradation reqvlre 



vlth FI greater than T 



are borderline c 
requiring use 01 
lyuboli. 























50 


Ccsaparlnc Soils at equal Liquid Uslt 
Tetajbnee* and Dry Strength Increase 
vlth Increasing Plasticity Indes 






























to 

1 

1 p 






:cl— 






:«~ 






T 
k 








.«.— 






-M 









•0 50 60 

LlflHD LOOT 
PLASTICITl CKAST 
classification of fine-grained sails 



axaarple OV-QC, ■ell-graded g 



-aand ■liturt with clay binder. (2) 






Table 2-3. Unified Soil Classification System 



U. S. Army Corps of Engineers 



2-7 



TM 5-818-1 / AFM 88-3, Chap. 7 



VOLUME 



WEIGHT 



I. 


I 






t 


1 




0» 

> 




GAS 


o 




> 

> 


' 


■ 




1 


1 


I 


I 






i 

> 




WATER 


* 

* 


* 


( 
- 

It 

> 


' 


1 






> 




W4 


' 


'///%% 




' • 



WATER CONTENT 


w 


= 


w, 


SPECIFIC GRAVITY 


G s 


= 


v s rw 


VOLUME OF SOLIDS 


Vs 


= 




VOLUME OF VOIDS 


Vv 


= 


v-v s 


VOID RATIO 


e 


r 


V v _ n 
V, " 1-n 


POROSITY 


n 


_ 


V v e 



1+e 



DEGREE OF SATURATION 



V w _ w G s 
V„ e 



UNIT WEIGHT OF WATER 
(FRESH WATER) 

DRY UNIT WEIGHT 



y w = -=■ = 62.4 PCF 



76 = 



%i _ 7m 

1 + w 



WET UNIT WEIGHT 



7m - y 



U. S. Army Corps of Engineers 



SUBMERGED (BOUYANT) UNIT WEIGHT y' = y m - y m = -^— y„ 



1 + e 



U. S. Army Corps of Engineers 



Figure 2-3. Weight-volume relationships. 



2-8 



TM 5-818-1 / AFM 88-3. Chap. 7 

Table 2-4. Descriptive Soil Names Used in Local Areas (L)and Names Widely Used 



Alluvium 

Argillaceous 
Bentonite 

(L) Boulder clay 
(L) Buckshot 

(L) Bull's liver 

Calcareous 

Caliche 

(L) Coquina 



Diatomaceous 
earth 



(L) Dirty land 

Disintegrated 
granite 

|L| Fat clay 

Fuller' a earth 



(L) Gumbo 



Ha rdpan 



Late ri tic soils 



Lean clay 
<L> Lime rock 



Calcareous silts and sandy-silty clays which are 
usually high in colloidal clay content, found in the 
semi a rid regions of the southwestern United States 
and North Africa. 

Deposits of mud, silt, and other material commonly 
found on the flat lands along the lower courses of 
streams. 

Soils which are predominantly clay or abounding in 
clays or clay-like materials. 

A clay of high plasticity formed by the decom- 
position of volcanic ash; it has high swelling 
characteristics. 

Another name, used widely in Canada and England, 
for glacial till. 

Clays of the southern and southwestern United 
States which, upon drying, crack into Bmall, hard 
lumps of more or less uniform size. 

This is a name used in some sections of the 
United States to describe an inorganic silt of 
slight plasticity. When saturated, it quakes like 
jelly from vibration or shock. 

Soils which contain an appreciable amount of 
calcium carbonate, usually from limestone. 

This term is widely used in construction to de- 
scribe deposits which contain various amounts of 
silt, clay, and sand cemented by calcium carbon- 
ate deposited by evaporation of groundwater, as 
found in France, North Africa, Texas, and other 
southwestern states. 

Consists essentially of marine shells which are 
held together by a small amount of calcium car- 
bonate to form a fairly hard rock. Coquina 
shells (and oyster shells) are widely used for 
granular stabilisation of soils along the Gulf 
Coast of the United States. 

Calcareous, rock-like material formed by secre- 
tions of corals and coralline algae. 

Composed essentially of the siliceous skeletons of 
diatoms (extremely small uni celled organisms). It 
is composed principally of silica, is white or light 
gray in color, and extremely porous. 

A slightly silty or clayey sand. 

Granular soil derived from advanced weathering 
and disintegration of granite rock. 

Fine, colloidal clay of high plasticity. 

Unusually highly plastic clays of sedimentary 
origin, white to brown in color. Used commercially 
to absorb fats and dyes. 

Peculiar, fine-grained, highly plastic silt-clay soils 
which become impervious and soapy, or waxy and 
sticky, when saturated. 

A general term used to describe a hard, cemented 
soil layer which does not soften when wet. Use of 
this term should be avoided since it implies a con- 
dition rather than a type of soil. 

Residual soils which are found in tropic regions. 
Many different soils are included in this category 
and they occur in many sections of the world. They 
are frequently red in color, and in their natural 
state have a granular structure with low plasticity 
and good drainability. When they are remolded in 
the presence of water, they often become plastic 
and clayey to the depth disturbed. 

Silty clays and clayey silts, generally of low to 
medium plasticity. 

A soft, friable, compact, cream-white, high- 
calcium limestone found in the southeastern United 
States which consists of coral and other marine re- 
mains which have been disintegrated by weathering. 

A general agricultural term which is applied most 
frequently to sandy-silty topsoils which contain a 
trace of clay, are easily worked, and are productive 
of plant life. 



Micaceous 
soils 



Muck (mud) 
Peat 



Muskeg 

(L) Red dog 
Rock flour 



Topsoil 

Tufa 
Tuff 

Varved clay 
Volcanic ash 



Silty soil of aeolian origin characterized by a 
loose, porous structure, and a natural vertical 
slope. It cover b extensive areas in North America 
(especially in the Mississippi Basin), Europe, and 
Asia (especially North Central Europe, Russia, 
and China). 

A soft, calcareous deposit mixed with clays, silts, 
and sands, often containing shells or organic re- 
mains. It is common in the Gulf Coast area of the 
United States. 

Soil which contains a sufficient amount of mica to 
give it distinctive appearance and characteristics. 

The very soft, slimy silt or organic silt which is 
frequently found on lake or river bottoms. 

A term which is frequently applied to fibrous, 
partially decayed organic matter or a soil which 
contains a Large proportion of such materials. 
Large and small deposits of peat occur in many 
areas and present many construction difficulties. 
Peat is extremely loose and compressible. 

Peat deposits found in northwestern Canada and 
Alaska. 

The residue from burned coal dumps. 

A fine-grained soil, usually sedimentary, of low 
plasticity and cohesion. Particles are usually in 
the lower range of silt sizes. At high moisture 
contents, it may become "quick" under the action 
of traffic. 

A thinly laminated rock-like material resulting 
from consolidation of clay under extreme pressure. 
Some shales revert to clay on exposure to air and 
moisture. 

A fan-shaped accumulation of mixed fragments of 
rock that have fallen, because of weathering, at or 
near the base of a cliff or steep mountainside. 

A general term applied to the top few inches of 
soil deposits. Topaoils usually contain consider- 
able organic matter and are productive of plant 
life. 

A loose, porous deposit of calcium carbonate which 
usually contains organic remains. 

A term applied to compacted deposits of the fine 
materials ejected from volcanoes, such as more or 
less cemented dust and cinders. Tuffs are more or 
less stratified and in various states of consolida- 
tion. They are prevalent in the Mediterranean 
area. 

A sedimentary deposit which consists of alternate 

thin (1/8 in. to 1/2 in.) layers of silt and clay. 

Uncemented volcanic debris, usually made up of 
particles less than 4 mm in diameter. Upon 
weathering, a volcanic clay of high compressibility 
is frequently formed. Some volcanic clays present 
unusually difficult construction problems, as do 
those in the area of Mexico City and along the 
eastern shores of the island of Hawaii. 



U. S. Army Corps of Engineers 



2-9 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 2-5. A Simplified Classification for Rocks 



u 

M 
U 

O 

OS 

3 
O 
%i 

a 

0* 


Color 


Principal 

Minerals 


Texture 


Very Coarse, 

Irregular 
Crystalline 


Coarse and 

Medium 
Crystalline 


Fine 

Crystalline 


Micro- 
crystalline 


Glassy 


Porous 
(Gas Openings) 


Fragmental 


Light 


Quartz and 
Feldspar 

Feldspar, 
Little or no 
Quartz 


Pegmatite 

Syenite 
Pegmatite 


Granite 
Syenite 


Aplite 


*j 

« 


Rhyolite 


Pitchstone 


Pumice 


i 

%> - 

c .— 

■-. « 

'—• •* 

•i 

«w '-v 

*- *J tl 

3 w — 

4-1 A 

— * u w 
«l (J M 

M 4J > 

O *- *-' 
O .Q 
-h n 
- u 
-^s 4J 
*l "« "3 

e 4i e 

-H .hi -.* 

*- B u 

:« 

< 


Trachyte 


Andes ite 


Intermediate 


Feldspar and 
Hornblende 


Diorite 
Pegmatite 


Diorite 


* 
« 

O 

Q 


Scoria or 

vesicular 

basalt 


Diabase 


Dark 


Augite and 
Feldspar 


Gabbro 
Pegmatite 


Gabb ro 




Augite, 

Hornblende 
Olivine 




Peridotite 







o 
at 

i 

u 


Group 


Grain Size 


Composition 


Name 


Clastic 


Appreciable quantity 

of grains more 
than 2-mm diameter 


Rounded pebbles in medium-grained matrix 


Conglomerate 


Angular coarse rock fragments, often quite variable 


Breccia 


More than 50% of 

grains are 0.06* to 

2.00-mm diameter 


Medium 
quartz 

grains 


Less than 10% of other minerals 


Siliceous sandstone 


Appreciable quantity of clay minerals 


Argillaceous sandstone 


Appreciable quantity of calcite 


Calcareous sandstone 


Appreciable quantity of iron oxide cement 


Ferruginous sandstone 


Over 251 feldspar; less than 751 quartz 


Arkose 


10-50% feldspar and darker minerals; 
30 to 40% quartz 


Grayvacke 


More than SOX of grains 
are 0.002-to 0.06-mm 
diameter 


Fine to very fine quartz grains with clay minerals 


Siltstone (if laminated, shale) 


Predominantly grains 
less than 0.002-mm 

diameter 


Microscopic 
clay minerals 


Less than 10% other minerals 


Shale (if not laminated, claystone) 


Appreciable calcite 


Calcareous shale 


Appreciable carbonaceous material 


Carbonaceous shale 


Appreciable iron oxide cement 


Ferruginous shale 


Organic 


Variable 


Calcite and fossils 


Fossiliferous limestone 


Carbonaceous material 


Bituminous coal 


Chemical 


Microcopic 


Calcite 


Limestone 


Dolomite 


Dolomite 


Quartz 


Chert, flint, etc. 


Iron compounds with quartz 


Iron 


Halite 


Rock salt 


Gypsum 


Rock gypsum 



o 

X 

u 

-C 

a. 



B 
n 


Fol iation 


Texture 


Coarse Crystalline & Banded 


Coarse Crystalline 


Medium Crystalline 


Fine to Microscopic Crystalline 


Foliated 


Gneiss 


( Sericite 
1 Mica 
Schist ( Talc 

J Chlorite 

' Hematite, etc. 


Phyllite 


Slate 


Non- 
Foliated 




Marble \ 
Quartzite f 
Serpentine / 
Soapstone ' 


Marble \ 
Quartzite f 
Serpentine ( 
Soapstone J 


Hornfels 

Anthractive Coal 
Marble \ 
Quartzite f a 
Serpentine ? 
Soapstone ) 



Variable grain size. 



U. S. Army Corps of Engineers 



2-10 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 2-6. Descriptive Criter for Rock 



1. Rock type 

a. Rock name (Generic) 

b. Hardness 

(1 ) Very soft: can be deformed by hand 

(2) Soft: can be scratched with a fingernail 

(3) Moderately hard: can be scratched easily with a knife 

(4) Hard: can be scratched with difficulty with a knife 

(5) Very hard: cannot be scratched with a knife 

c. Degree of weathering 

(1 ) Unweathered: no evidence of any mechanical or chemical alteration. 

(2) Slightly weathered: slight discoloration on surface, slight alteration along discontinuities, less than 10 
percent of the rock volume altered, and strength substantially unaffected. 

(3) Moderately weathered: discoloring evident, surface pitted and altered with alteration penetrating well 
below rock surfaces, weathering "halos" evident, 10 to 50 percent of the rock altered, and strength 

noticeably less than fresh rock. 

(4) Highly weathered: entire mass discolored, alteration in nearly all of the rock with some pockets of slightly 
weathered rock noticeable, some minerals leached away, and only a fraction of original strength (with wet strength usually 
lower than dry strength) retained. 

(5) Decomposed: rock reduced to a soil with relect rock texture (saprolite) and generally molded and 
crumbled by hand. 

d. Lithology (macro description of mineral components). Use standard adjectives, such as shaly, sandy, silty, 
and calcareous. Note inclusions, concretions, nodules, etc. 

e. Texture and grain size 
(1) Sedimentary rocks 

Tfixtnrfi Grain niamfitfir mm Partirlfi Nams Rnrk Nams 

Conglomerate 



Grain niamptpr 


mm 


Partirlfi Namp 


<-80 




Cobble 


5 -.80 




Gravel 


2-5 






0.4-2 




Sand 


0.1 -0.4 







Coarse grained 

Medium grained 0.4 - 2 Sand Sandstone 

Fine grained 

Use clay-sand texture to describe conglomerate matrix. 

(Continued) (Sheet 1 of 3) 



2-11 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 2-6. Descriptive Criteria for Rock-Continued 
Texture Grain Diameter, mm Particle Name Rock Name 



Very fine grained >-0.1 Clay, Silt Shale, Claystone, 

Siltstone, 
Limestone 

(2) Igneous and metamorphic rocks 

Texture Grain Diameter, mm 

Coarse grained > - 5 

Medium grained 1 - 5 

Fine grained 0.1 - 1 

Aphanite < 0.1 

(3) Textural adjectives. Use simple standard textural adjectives such as porphyritic, vesicular, pegmatitic, granular, 
and grains well developed. Do not use sophisticated terms such as holohyaline, hypidiomorphic granular, crystaloblastic, 
and cataclastic. 

2. Rock structure 

a. Bedding 

(1) Massive: >3 ft thick 

(2) Thick bedded: beds from 1 to 3 ft thick 

(3) Medium bedded: beds from 4 in. to 1 ft thick 

(4) Thin bedded: beds less than 4 in. thick 

b. Degree of fracturing (jointing) 

(1 ) Unfractured: fracture spacing greater than 6 ft 

(2) Slightly fractured: fracture spacing from 3 to 6 ft 

(3) Moderately fractured: fracture spacing from 1 to 3 ft 

(4) Highly fractured: fracture spacing from 4 in. to 1 ft 

(5) Intensely fractured: fracture spacing less than 4 in. 

c. Shape of rock blocks 

(1) Blocky: nearly equidimensional 

(2) Elongated: rod-like 

(3) Tabular: flat or bladed 

(Continued) 

(Sheet 2 of 3) 

2-12 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 2-6. Descriptive Criteria for Rock-Continued 



Discontinuities 

a. Joints 

(1) Type: bedding, cleavage, foliation, schistosity, extension 

(2) Separations: open or closed, how far open 

(3) Character of surface: smooth or rough; if rough, how much relief, average asperity angle 

(4) Weathering of clay products between surfaces 

b. Faults and shear zones 

(1) Single plane or zone: how thick 

(2) Character of sheared materials in zone 

(3) Direction of movement, slickensides 

(4) Clay fillings 

c. Solution, cavities, and voids 

(1) Size 

(2) Shape: planar, irregular, etc. 

(3) Orientation (if applicable): developed along joints, bedding planes, at intersections of joints and bedding 

(4) Filling: percentage of void volume and type of filling material (e.g. sand, silt, clay, etc.). 



planes, etc. 



U. S. Army Corps of Engineers 

ment of texture or structural features in any type of rock; 
e.g., cleavage in slate or schistosity in a metamorphic 
rock. The term is most commonly applied to 
metamorphic rock. 



(Sheet 3 of 3) 



Referring to l figure 2" : 4~l with a core advance of 60 inches 
and a sum of intact pieces, 4 inches or larger, of 34 
inches, the RQD is computed as: 
34 
RQD =60 = 0.57 



b. Samples that are tested in the laboratory 
(termed "intact" samples) represent the upper limit of 
strength and stress-strain characteristics of the rock and 
may not be representative of the mass behavior of the 
rock. Coring causes cracks, fissures, and weak planes 
to open, often resulting in a recovery of many rock 
fragments of varying length for any core barrel advance. 
Only samples (intact pieces) surviving coring and having 
a length/diameter ratio of 2 to 2.5 are tested. Rock 
Quality Designation (RQD) is an index or measure of the 
quality of the rock mass. RQD is defined as: 

L Lengths of intact 

RQD = pieces >_4 in. long 

Length of core advance 



Also shown i n! figure 2 : 4| is a qualitative rating of the rock 
mass in terms of RQD. RQD depends on the drilling 
technique, which may induce fracture as well as rock 
discontinuities. Fresh drilling-induced fractures may be 
identified by careful inspection of the recovered sample. 
2-7. Shales. 

a. Depending on climatic, geologic, and 
exposure conditions, shale may behave as either a rock 
or soil but must always be handled and stored as though 
it is soil. For these reasons, shale is considered 
separately from either soil or rock. Shale is a fine-grained 
sedimentary rock composed essentially of compressed 



2-13 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 2- 7. Engineering Classification of Intact Rock 



On basis of strengh, o u]t : 



Class 
A 
B 
C 

D 
E 



Description 
Very high strength 
High strength 
Medium strength 
Low strength 
Very low strength 



Uniaxial Compressive 

Over 32,000 

16,000-32,000 

8,000-16,000 

4,000 - 8,000 

Less than 4,000 



On basis of modulus ratio, E t /o U | t : 



Class 
H 

M 
L 



Description 
High modulus ratio 
Average (medium) ratio 
Low modulus ratio 



Uniaxial Compressive 

Over 500 

200-500 

Less than 200 



a Rocks are classified by both strength and modulus ratio, such as AM, BL, BH, and CM. 

b Modulus ratio = E t /a u i t , where E t = tangent modulus at 50 percent ultimate strength and a u it = uniaxial compressive 

strength. 



and/or cemented clay particles. It is usually laminated 
from the general parallel orientation of the clay particles 
as distinct from claystone, siltstone, or mudstone, which 
are indurated deposits of random particle orientation. The 
terms "argillaceous rock" and "mudrock" are also used to 
describe this type of rock. Shale is the predominate 
sedimentary rock in the earth's crust. 

b. Shale may be grouped as compaction 
shale, and cemented (rock) shale. Compaction shale is a 
transition material from soil to rock and can be excavated 
with usual large excavation equipment. Cemented shale 
generally requires blasting. Compaction shales have 
been formed by consolidation pressure and very little 
cementing action. Cemented shales are formed by a 
combination of cementing and consolidation pressure. 
They tend to ring when struck by a hammer, do not slake 
in water, and have the general characteristics of 



(Courtesy of K. G. Stagg and O. C. Zienkiewiez, Rock 
Mechanics in Engineering Practice, 1968, pp 4-5. 
Reprinted by permission of John Wiley & Sons, Inc, New 
York.) 

good rock. Compaction shales, being of an interme- diate 
quality, will generally soften and expand upon exposure 
to weathering when excavations are opened. 

c. Dry unit weight of shale may range from 
about 80 pounds per cubic foot for poor-quality 
compaction shale to 160 pounds per cubic foot for high- 
quality cemented shale. Shale may have the appearance 
of sound rock on excavation but will often deteriorate, 
during or after plac ement in a f ill, into weak clay or silt, of 
low shear strenqth. rFiqure 2-5l mav be used as a guide in 
classifying shale for foundation use. 

d. Compaction shales may swell for years after 
a structure is completed and require special studies 
whenever found in subgrade or excavated slopes. The 
predicted behavior of shales cannot be based sofely 
upon laboratory tests and must recognize local 
experiences. 



2-14 



TM 5-818-1 / AFM 88-3, Chap. 7 



CORE • 
RECOVERY 
INCHES 



10 

2 

6 

3 
4 

5 
3 

4 



4 
2 
5 

TOTAL = 50 INCHES 










ES3 



MODIFIED 

CORE RECOVERY 

INCHES 



10 



4 

5 



ROCK QUALITY RATING 



RQD 


DESCRIPTION 


0-25 


VERY POOR 


25-50 


POOR 


50-75 


FAIR 


75-90 


GOOD 


90-100 


EXCELLENT 



CORE RECOVERY - 50/60 - 83% 
RQD = 34/60 - 57% 



TOTAL -34 INCHES 



(Courtesy of K. G. Stagg and O. C. Zienkiewiez, Rock 

Mechanics in Engineering Practice. 1968.pl5. Reprintedby 

permission of John Wiley & Sons, Inc. New York.) 



Figure 2-4. Modified core recovery as in index of rock quality. 



2-15 



TM 5-818-1 / AFM 88-3, Chap. 7 



ARGILLACEOUS MATERIALS 



s uo <250PS>] 

As u >0.6 s u 
Aw>l% 



L. 



_I 



s u0 > 250 PSI (IB TSF) 
Aw<l% 



CLAY 



MUDSTONE* 



I 

MEDIUM 

TO 

SOFT 

t^lHR 



f 

STIFF 



t M <lDAY 



f 

HARD 
(CLAY-SHALE) 



1 M >1DAY 



t 

CLAYSTONE 



SILTSTONE 
•(SHALE IF FISSILE) 



s = UNDRAINED SHEAR STRENGTH AT NATURAL WATER CONTEUT 
u0 (Q TEST USING 50 PSI CHAMBER PRESSURE) 

As u = STRENGTH LOSS AFTER SOFTENING TO EQUILIBRIUM WATER CONTENT 

Aw = CHANGE IN WATER CONTENT AFTER SOFTENING 

U = TIME OF SOFTENING FOR LOSS OF 50% OF s 

3U UO 

a. ENGINEERING CLASSIFICATION OF ARGILLACEOUS MATERIALS 







AMOUNT OF SLAKING 


a 




LL = LIQUID LIMIT 


VERY LOW 


LOW 


MEDIUM 


HIGH 


VERY HIGH 


w, = MAX. WATER CONTENT DUE 
TO SLAKING 

li = LIQUIDITY INDEX 


VL 
LL<20 


L 

LL 
BETWEEN 


M 

LL 
BETWEEN 


H 

LL 
BETWEEN 


VH 
LL>140 


Ali 


= CHANGE IN LIQUIDITY INOEX 




20ft 50 


50&90 


90 & 140 






z 


SLOW, S 


VL 


L 


M 


H 


VH 




t4 

5 


IS* 

ce 

UJ 

Si 


Ali <0.75 


S 


S 


S 


S 


S 




FAST, F 


VL 


L 


M 


H 


VH 




CO 

u_ 

O 
UJ 


« * 

"i °° 


0.75<Ali<1.25 


F 


F 


F 


F 


F 




VERY FAST, VF 


VL 


L 


M 


H 


VH 




ac 


< t 


Ali > 1.25 


VF 


VF 


VF 


VF 


VF 



a WATER CONTENT AFTER SLAKING EQUALS LL. 

b. CLASSIFICATION IN TERMS OF SLAKING CHARACTERISTICS 

(Courtesy of N. R. Morgenstern and K. D. Eigenbrod, "Classification of Argillaceous Soils and Rock, " Journal, 
Geotechnical Division, Vol lOO. No. GTIO, 1974. pp IW-1ISS. Reprinted by permission of American Society of Civil 

Engineers. New York.) 



Figure 2-5. Classification of shales. 
2-16 



TM5-818-1/AFM 88-3, Chap. 7 



CHAPTER 3 



ENGINEERING PROPERTIES OF SOIL AND ROCK 



3-1. Scope. This chapter considers engineering 
properties of soil and rock useful in designing 
foundations under stati c loading. Dynamic properties are 
discussed ir l chapterT| 7. 

a. Correlations. Tables and charts based on 
easily determined index properties are useful for rough 
estimating or confirming design parameters. Testing 
procedures employed by different soil laboratories have 
influenced correlations presented to an unknown degree, 
and the scatter of data is usually substantial; caution 
should, therefore, be exercised in using correlation 
values. Undisturbed soil testing, either laboratory or field, 
or both, should be used for final design of major 
foundations. On smaller projects, an economic analysis 
should determine if a complete soil exploration/laboratory 
testing program is justified in lieu of a conservative 
design based on correlation data. Complex subsurface 
conditions may not permit a decision on solely economic 
grounds. 

b. Engineering properties. Properties of 
particular interest to the foundation engineer include- 

(1) Compaction. 

(2) Permeability. 

(3) Consolidation-swell. 

(4) Shear strength. 

(5) Stress-strain modulus (modulus of 
elasticity) and Poisson's ratio. 

3-2. Compaction characteristics of soils. 
The density at which a soil can be placed as fill or backfill 
depends on the placement water content and the 
compaction effort. The Modified Compaction Test (CE 
55) or comparable commercial standards will be used as 
a basis for contro l. The CE 5 5 test is described in TM 5- 
824-2/AFM 88-6, Chapter 2. 1 (See app A for references.) 
Other compaction efforts that may be occasionally used 
for special projects include- 

a. Standard compaction test: Three layers at 
25 blows per layer Hammer = 5.5 pounds with 12-inch 
drop 

b. Fifteen-blow compaction test: 
Three layers at 15 blows per layer 
Hammer = 5.5 pounds with 12-inch drop 

The results of the CE 55 t est are re presented by 
compaction curves, as shown i h figure 3"^ , in which the 
water content is plotted versus compacted dry density. 
The ordinate of the peak of the curve is the maximum dry 
density, and the abscissa is the optimum water content 



W opt . I Table 3-il presents typical engineering properties 
of compacted soils; see footnote for compacted effort 
that applies. 

3-3. Density of cohesionless soils. 

a. Relative density of cohesionless soils has a 
considerable influence on the angle of internal friction, 
allowable bearing capacity, and settlement of footings. 
An example of the relationship between relative density 
and in situ dry d ensities may be conveniently determined 
from lfigure 3-2.1 Methods for determining in situ densities 
or relative densities of sands in the field are discussed in 

I chapter 41 

b. The approximate relationship amo ng the 
ang le of friction, +, DR, and unit weight is shown in | figurel 

13-31 and be tween the coefficient of uniformity, Cu, and 
void ratio, in lfigure S^ 

c. The relative compaction of a soil is defined 
as 

RC = — ~- ~ ----- x 1 0O(percent) (3-1 ) 

' max (lab) 

where yf ie id = dry density in field and y m ax(iat>) = maximum 
dry density obtained in the laboratory. For soils where 
100 percent relative density is approximately the same as 
100 percent relative compaction based on CE 55, the 
relative compaction and the relative density are related 
by the following empirical equation: 

RC = 80 + 0.2D R (D R > 40 percent) (3-2) 

3-4. Permeability. 

a. Darcy's law. The laminar flow of water 
through soils is governed by Darcy's law: 

q = kiA (3-3) 
where 

q = seepage quantity (in any time unit consistent 

with k) 
k = coefficient of permeability (units of velocity) 
i = h/L = hydraulic gradient or head loss, h, 

across the flow path of length, L 
A = cross-sectional area of flow 

b. Permeability of soil. The permeability - 
depends primarily on the size and shape of the soil 
grains, void ratio, shape and arrangement of voids, 
degree of saturation, and temperature. Permeability is 
determined in the laboratory by measuring the rate of 
flow of wa- 



3-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



ter through a specimen under known hydraulic gradient, i. 
Typical permeability values, empirical relationships, and 
methods for obtaining the coefficient of permeability are 
shown in Ifigure 3-571 Field pumping tests are the most 
reliable mean s of determ ining the permeability of natural 
soil deposits kpara 4-5l Permeability obtained in this 



manner is the permeability in a horizontal direction. The 
vertical permeability of natural soil deposits is affected by 
stratification and is usually much lower than the 
horizontal permeability. 



\£U 








"1 ! ! 

V NOTE: DETERMINATION OF OPTIMUM 










\ WATER CONTENT AND MAXIMUM 










\ DRY DENSITY SHOWN FOR LEAN 










\ CLAY ONLY. 










\ SAMPLES COMPACTED USING CE 55 


115 








\ ^-^MPArTI^N EFFORT "~L^5S 1B:| - 










\ CATION DATA FOR SAMPLES SHOWN 












IN FIGURE 2-2 




110 








\\i. 
























MAXIMUM DRY DENSITY 


\"^ 










= 107 LB/CU FT 












OPTIMUM WATER 




_j& 






105 




CONTENT - 18.0% 
















k Vo" 














I 


\ \ * 














/*\ 


\ \% 














f \ 


\ \ ** 










, 


\ 


\ \ Q. 




U. 






J 


\ 


\ > ^ 




u 100 






1 


\ 


LEAN CLAY 


\o 




Q. 






^~~ t 


\ 






>" 






/ 


\ 




\ <° 




1- 






/ 


> 


k 


V 




1/1 

z 
u 


UNIFORM FINE 


/ 




\ 






SAND 


1 




\ 






Q 










SILT 






> 95 
X 






























Q 


















I 


v^ 










90 


I 


MAX DRY DENSITY \ r 














i-^^^^^ffi 












Ul 


COMPACTION ft (_ >. \tj. 












< 


CURVE. / Z \ \* 












bj 


+*/ lii \ \^. 












K 


/ *-' \\b 












O 


/ °' w 












z 










85 




/ K A\£v 
/ Ul Wi 
















/ FA i \..l.«>* x 




>- 
1- 


/ <l \v 












<o 


/ * \\ 












z 

UJ 
Q 


/ si V 

§l 










80 


> 


El 










Q 














LUAT^e* ^AMTPUT ikJj—O 






75 


Tf 


« i cr* ^,un i c 


■* i in^r* 


tAot " m 



10 15 20 25 

WATER CONTENT, PERCENT DRY WEIGHT 



30 



35 



U. S. Army Corps of Engineers 



Figure 3- 1 . Typical CE 55 compaction test data. 
3-2 



TM 5-818-1 / AFM U-3. Chap. 7 

Table 3- 1 . Typical Engineering Properties of Compacted Materials 

' * Typical Value of "" ~~~ 

Cowpresslon Typical Strength Characteristics 

Ranee of Range of At 2.5 At T.2 Range of 

Maximum Optimal ksf ksf Cohesion t (Effective Typical Subgrade 

Dry Unit _ Water, (20 pai) (50 pal) {As Cob- Cohesion Stress Coefficient of Modulus 

Weight, Content Percent of Original pacted) (Saturated) Qivelope) Permeability Range of k 

pcf Percent Height psf psf deg ft/In CBR Values lb/cu in. 

125-135 11-8 0.3 0.6 >38 5 " in"' UO-80 300-500 

115-125 1»-11 0.U 0.9 >3T 10" 1 30-60 250-tOO 

120-135 12-8 0.5 1.1 >3!< >10. 20-60 100-»00 

115-130 1U-9 0.7 1.6 >31 'l, " 7 20-«0 100-300 

110-130 16-9 0.6 1.2 38 >10* 3 20-10 200-300 

100-120 21-12 0.8 l.k 3T >10 -3 10-k0 200-300 

110-125 16-11 0.8 1.6 1050 120 3" 5 » 10" 5 10-lt0 100-300 

110-130 15-11 0.8 l.li 1050 300 33 2» 10" 6 

105-125 19-11 1.1 2-2 1550 230 31 5 » 10~ 7 5-20 100-300 

95-120 2U-12 0.9 1-7 IkOO 190 32 lo" 5 15 or less 100-200 

100-120 22-12 1.0 2.2 1350 I460 32 5 " 10~ 7 100-200 

95-120 2U-12 1.3 2-5 1800 270 28 10~ 7 15 or less 50-200 

80-100 33-?l 5 or less 50-100 

75-95 k0-2l 2.0 3.8 1500 1)20 25 5 » 10~ 7 10 or less 50-100 

80-105 36-19 2.6 3.9 2150 230 19 10" 7 15 or less 50-150 

75-100 U5-21 5 or less 25-100 



Symbol 


Soil Type 


GW 


Veil graded clean grav- 
els, gravel- sand 
mixtures 


GP 


Poorly graded clean 
gravels, gravel-Band 

nix 


CM 


Silty gravels, poorly 
graded gravel-sand- 

silt 


GC 


Clayey gravels, poorly 
graded gravel -sand- 
clay 


SW 


Veil graded clean sands, 
gravelly sands 


SP 


Poorly graded clean 
sands, sand-gravel 
nix 


SM 


Siity sands, poorly 
graded sand-silt mix 


SM-SC 


Sand-silt clay nix. with 
slightly plastic 
fines 


SC 


Clayey sands, poorly 
graded sand-clay mix 


ML 


Inorganic silts and 
clayey silts 


ML-CL 


Mixture of inorganic 
silt and clay 


CL 


Inorganic clays of lov 
to med. plasticity 


OL 


Organic silts and silt- 
clays, lov plasticity 


MH 


Inorganic clayey ailta, 
elastic silts 


CH 


Inorganic clays of high 
plasticity 


OH 


Organic clays and silty 
clays 



Notes 1. All properties are for condition of "standard Proctor" maximum density, except values of k and CBH which 
are for CE55 .maximum density. 

2. Typical strength characteristics are for effective strength envelopes and are obtained from ISBR data. 

3. Compression values are for vertical loading with complete lateral confinement. 

*. 4. (>) indicates that typical property is greater than the value shown. ( ) indicates insufficient data available for an 

estimate. 



(NAVFACDM-7) 



3-3 



TM 5-818-1 / AFM 88-3. Chap. 7 



150 



140 



Z 130 



^ 



120 



U. 
U 
0. 

> 

H 
VI 

Z 

w 

Q 

>- 

ir 
o 



110 



100 



90 



TO USE: 


1 1 

PLOT y Hfl =99.0, y MA x = 119 PCF, AS 
SHOWN AND CONNECT WITH STRAIGHT 






LINE yj = 1 10 
y A - 114.4 PCF 


1 PCF FOR O r 
" FOR D, = 80% 


= 60% 
ETC. 




























^"— EXAMPLE 















150 



140 



130 



120 
U. 
O 

a. 

> 

110 •- 
in 

z 

UJ 

□ 

>- 
a. 

100 o 



20 



40 60 ao 

RELATIVE DENSITY, D r PERCENT 



100 



90 



Figure 3-2. Relation between relative density and dry density (scaled to plot as a straight line). 



0.55 



0.5 



0.45 



POROSITY, n (FOR G - 2.681 
0.4 0.35 0.3 0.25 



0.2 



0.15 



-i i i 1 1 1 r 

VOID RATIO, e IFOR G - 2.68) 
1.2 1.1 1.0 0.9 0.8 0.75 0.7 0.65 0.6 0.55 0.5 0,46 0.4 0.35 0.3 0.25 0.2 0.15 



45 



40- 



oc 



(- 35 — 

o 

E 



< 
z 
cc 
uj 30 

z 

u. 

o 

UJ 

-I 

o 

Z 25 — 



20 



t — i i—m — i — i — i — i — i — r 



MATERIAL TYPE 




^SUANDSP 
IN THIS RANGE 



# OBTAINED FROM 
EFFECTIVE STRESS 
FAILURE ENVELOPES 

APPROXIMATE CORRELATION 
IS FOR COHESIONLESS 
MATERIALS WITHOUT 
PLASTIC FINES 



A 



_L 



75 80 90 100 110 120 

DRY UNIT WEIGHT iy D ) PCF 



130 



140 



150 



(NAVFAC DM-7) 



Figure 3-3. Angle of friction versus dry density for coarse-grained soils. 



3-4 



TM 5-818-1 / AFM 88-3, Chap. 7 



c. Permeability of rock. Intact rock is 
generally impermeable, but completely intact rock 
masses rarely occur. The permeability of rock masses is 
controlled by discontinuities (fissures, joints, cracks, 
etc.), and flow may be either laminar (Darcy's law 
applies) or turbulent, depending on the hydraulic 
gradient, size of flow path, channel roughness, and other 
factors. Methods for d etermining the in situ permeability 
of rock are presented in lchapterU 

3-5. Consolidation. Consolidation is a time-de- 
pendent phenomenon, which relates change that occurs 
in the soil mass to the applied load. 

a. Consolidation test data. Consolidation or 
one-dimensional compression tests are made in 
accordan ce with accepted standards. Results of tests 
tfiq 3-6)1 are presented in terms of time-consolidation 
curves and pressure-void ratio curves. The relationship 
between void ratio and effective vertic al stress, p , is 
shown on a semilogarithmic diagram in Ifigure 3-6.1 The 



test results may also be plotted as change in volume 
versus effective vertical stress. Typical examples of 
pressure - void ratio curves for insensitive and sensitive, 
normally loaded cl ays, and preconsolidated clays are 
shown in ! figure 3^71 

b. Preconsolidation pressure. The 

preconsolidation stress, pc, is the maximum effective 
stress to which the soil has been exposed and may 
result from loading or drying. Geological evidence of 
past loadings should be used to estimate the order of 
magnitude of preconsolidation stresses before laboratory 
tests are performed. The Casagrande method of 
obtaining the preconsolid ation pres sure from 
consolidation tests is shown in lfigure 3-7.1 Determining 
the point of greatest curvature 



1.4 



o i.o — 



< 



O 
> 

Z 
D 

z 

X 

< 

2 



o.e — 




0.6 — 



0.4 



0.2 



< 



O 
> 

z 
z 



o.e 



0.6 



0.4 



? 0.2 




I** - 



2 3 4 6 10 15 

COEFFICIENT OF UNIFORMITY. C 



NOTE: THE MINIMUM VOID RATIOS WERE OBTAINED FROM SIMPLE SHEAR TESTS. CURVES ARE 
ONLY VALID FOR CLEAN SANDS WITH NORMAL TO MODERATELY SKEWED GRAIN-SIZED 
DISTRIBUTIONS. 

(Modified from ASTM STP 523 (pp 98- 1 12). Copyright ASTM, 1916 Race St., 
Philadelphia, PA. 19103. Reprinted/adapted with permission.) 

Figure 3-4. Generalized curves for estimating e max , and e min from gradational and particle shape characteristics. 



3-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



PUMPING RATE ~ 
GNOUND SURFACE — - » 



-pi ,;^ 



DRAWDOWN CURVE 

TEST WELL 




OBSERVATION WELLS 
Ofi PIEZOMETERS 



GRAVITY FLO* (SHOWN ABOVE! 



ARTESIAN FLOW 



k = ■ 



■ lh ,'- k l'' 

O 



-f»-i. 



21011), - h ( l 

WHERE D IS THICKNESS OF ARTESIAN AQUIFER 

FIELD PUMPING TESTS 




KAt 
WHERE 1= TIME 

CONSTANT -HEAD 
PERMEAaETER 



At h 



WHERE t= TIME 



FALLING -HEAD 
PERMEAMETER 



10.000 



8 5,000 

a 
o 



J 

5 1,000 



I 

s 

at 

hi 
o 



soo 



100 

oos 

























,? 


J 
























i °°~ 




J 


: 


' 




























r< 




s 
























A 


^ 




























u. 


fX 












PERMEABILITY OF 






















MISS RIVE* ALLUVIAL 

SANDS BASCO ON FIELD 




*»>y 
















































/ 

/ , 

Y / 






FOR UNIFORM SANDS 
IN A LOOSE STATE 

LIMITED TO 












y 










u - ™ w i« * "™ 












/ 
























/ 
































j 
































t i 
































































f tat 
































































f/yf AFPROXRAATE RELATIONSHIP BETWEEN 
y6j COEFFICIENT OF PERMEABILITY AND 
yf OKA IN SIZE FOR UNIFORM. CLEAN SANDS 




r 1 1 1 II 1 III 1 



0.1 



0.4 
Oie SIZE IN MM 



1.0 



2.0 



COMPUTATION OP PERMEABILITY FROM GRAIN SIZE 



PERMEABILITY CHMAC TERKTICS OF IOIU 



Coefficient of Permeability, « 


Relative 
I* ratability 


Soil Type 


Method of Dt 






cn/eee 


ft/aln 


ft/yr 


terml nation 


10 


20 


10.; i 10^ 
1.0? x 10' 


High 


Clean grave la 


i. 

K 
-. It 

"A 






• ? 


? 


1 


2 




Coarse aanda 






V* 

g. 

«H M 


1000 x 10 ^ 


0.2 


io, 500 




Medina eendj 




Z 


100 x 10 


0.02 


1,050 




Fine aanda and aand 




is 
a. 


5 • 








NedlUB 


and gravel mix- 




*■ 


*t 


-It 








tures 


r* 




Sf 


10 X 10 


0.002 


105 




Very fine sand 






8 1 














3 JS 


fi 






1 x 10 


2 x 10 


10.5 




Sllty aanda. 


sa 


i 

E 

i. 














organic silts 






o.i x io 7 

0.01 X 10 


0.2 x 10 I 
0.02 x 10" 


1.05 
0.105 


Lov 


Sllta, glacial till 
Silty clay 




















I 

ip 


^ 
















♦; 


100 X 10 " 9 


200 x 10" 9 


105 x IO" 1 


Practically 


"Impervious" soils. 




*-* 

£ 


a 


pa- 


10 X 10" 9 


20 x 10" 9 


10.5 x 10 


ljapervlous 


e.g., honogeneous 




H 


s 2 - 

5 •* ^ 


1 x 10" 9 


1 x 10~ 9 






clays belov tone 






r. 


1.05 x 10 




of weathering 






2 


















Soi 
o o -3 



U. S. Army Corps of Engineers 

Figure 3-5. A summary of soil permeabilities and methods of determination. 



3-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



STANDPIPE 



Va 



£ 



DIAL GAGE 



LOADING PISTON 



POROUS STONE 



$$^SOIL SPECIMENjSSS?*" 



POROUS STONE 



S$^VV\\VV6asev\VVVV^V<^ 



■KING 



CONSOLIDATION APPARATUS 




I 


















; 


I 


j 




!il 




' 






II 














i 












: ; 


i 








PRESSURE 








I 












r 


I 1 ! 








1 6.M ' 










_rj% 




















2 VOO 


•■ ! 


















I 






_"] ' fir 










| | 






I 
















I 








! il 








i 




I I 




















[ 






> 


'"! !|! 






i 




i 


i 










-f?>l i j 










i . 






' 


! li 








II 




! 1 

! i 


' i 














i 






' 
















' ! 


i 










■ 


i 


I 


! ' 

! i ' ! 


. 










'M 



* "2 5 



TIME IN MINUTES 

CLAYS 



TIME IN MINUTES 

SANOS 



EXAMPLES OF LABORATORY CONSOLIDATION CURVES 







rr 

j 
1 


-4-- 


\ 








T" 














Ifi 


i 















'O^o'O 


- |o 8, ' i 












li!i 


1 


c < 


e <- p l 








s|l!l: 




iog 10 p z - log 10 □, 


^^^. 


t 

- 1 . i 


: ... „- 









l\ i : 
i \i: 




^^^^ 






J-iiJ 


'. 

^ 
















rrrt- 



0.5 1.0 

PRESSURE IN TONS 5Q FT (p| 

CLAYS 



SO 10 



j j 


i 
j 




!h 




Mil! 


■ 1 1 

: 1 ' 








j i ■! 




0.49- 0.46 

1 iog ia io-loi to 




1 




; 


I'll 

1 1 ; i 




c t = 


03 


1 
| 




! ! 


1 


— 4 


IH 

— i . f. 


e, 






1 
1 












«■! 






! 


i 

1 1 










1 ' 


■ 




I 


l!i 




— 






i ! ;: 








I : 
1 ' 








[ ! ! 



OS i 



PRESSURE IN TONS/SO FT (p) 

SANOS 



EXAMPLES OF LABORATORY PRESSURE - VOID RATIO CURVES 



U. S. Army Corps of Engineers 



Figure 3-6. Examples of laboratory consolidation test data. 



3-7 



TM 5-818-1 / AFM 88-3, Chap. 7 




- LABORATORY P-C CURVE 
OH -UNDISTURBED" 

= field p-e curve 

- OVERBURDEN PRESSURE 
PRESSURE AFTER 
EXCAVATION 

= PRECONSOLIuATION 
PRESSURE 



PRESSURE HOC SCALE) 

IHSEMITIVE NCWHALLr LOA0ED CLAV 




PRESSURE (LOGSCALEl 

SCNSOiVE MORHaLLV LOaMO CLAY 





~--<-"^ 








■i 
% 


\ 
\ 
\ \ 
\\ 
\\ 
\ \ 



PRESSURE (LOG SCALE! 

PMCOMSCLIMTED CLAT 



TYPICAL LA50RAT0RY PRESSURE - VOID RATIO CURVES AND METHODS FOR CONSTRUCTING FIELD CONSOLIDATION CURVES 



~— -'^^^ ^^- LABOR* TO* 


P-* CURVE 










S.* 




B 








v^ 




1 


1 


TO DETERMINE PPECONSOLIDATlON 






~~~~^o 




a / 

2 \ 


PRESSURE P c 






\f 






i determine point of minimum 








/a 


RAOIUS, POINT A 








/ a 




1 DRAR A HORIZONTAL LINE B, 












THOOUOM POIMT A 
























LABORATORY PRESSURE - VOID 












RATIO CURVE AT POINT A 












a ORARALINE.D BISECTING THE 












ANGLE FrMMCD BV LINES ,B. AND C 












S EXTEND THE LOWER OB STRAIGHT 












LINE PORTION OF THE LABORATORY 












PRESSURE - VOID RATIO CURVE UNTIL 












IT INTERSECTS LINE, THE POINT 












OF INTERSECTION KPRCSCNTS THE 












PRECONSOLIOaTiON PRESSURE. Pc ■ 













PRESSURE lLOGSCALI 





ill' 


| III ! 1 




1 1 ! M M 


T 








r initial I 

COMKH.WATKIM 1 


, 


*"-J INITIAL UAL RFAOMG ' 

. | I| 


t 












, 




"^ « PSIMAffV COVSOUPA TK» 

i , 4 , . , 4 . 


'l 










LAKMATORV T<«* -CONSOLIDATION CuiTVE | 












_-- fOft A CIVfN LOAD INCREMENT 












1 . . i . _. . i_^_ 








^^F— u i sw 






PHIUAMl COHtOLIpATIOH — 






I • 




. . . 4 , 




' I^NJ' 


. , . .+ t ^_ 






* 




. / CONSOLIDATION 


^^^ ♦ ; + * ^+ + ■ 


HEIGHT OF SAMPLE " 


W 


1*00 IN. 




c «TrT s Hit 5 u 


- 4 


LOG OF 1 CVCLE - 1 


T 1B h 


' •■'■(rHr 


1 


1 ! i i hi 


c " ■» 


" 








. ; i i 


| 



TIME IN MINUTES 



GRAPHICAL CONSTRUCTION OF PRECONSOLIDATION PRESSURE 



GRAPHICAL CONSTRUCTION FOR DETERMINING PRIMARY CONSOLIDATION 
AND CALCULATION OF COEFFICIENT OF CONSOLIDATION 



Figure 3- 7. Analyses of consolidation test data. 



3-8 



TM 5-818-1 / AFM 88-3, Chap. 7 



requires care and judgment. Sometimes it is better to 
estimate two positions of this point-one as small as likely, 
and the other as large as plausible, consistent with the 
data-and to repeat the construction for both cases. The 
result will be a range of preconsolidation stresses. 
Because the determination of pc involves some 
inevitable inaccuracy, the range of possible values may 
be more useful than a single estimate which falls 
somewhere in the possible range. The higher the quality 
of the test specimen, the smaller is the range of possible 
pc values. Approximate val ues of preconsolidation 
pres sure may be estimated fron i Ifigure 3-8 br 3-9~IITablel 
1 3-2 1 can be used to obtain gr oss estimates of site 
preconsolidation. This table an dlfiaures 3-8I and 3-9I 
should be applied before consolidation tests are 
performed to assure test loads sufficiently high to define 
the virgin compression portion of e-log p plots. 

c. Compression index. The slope of the virgin 
compression curve is the compression index C c , defined 
in I figure 3^67! Compressi on index c orrelations for 
approximations are given in table 3-3.1 When volume 
change is expressed as vertical strain instead of change 
in void ratio, the slope of the virgin compression part 



of the £ versus log p curve is the compression ratio, CR, 
defined as 



CR 



Ae 
log Q2. 
Pi' 



C c 
1 +e 



(3-4) 



where & is the change in vertical strain corresponding to 
a change in effective stress from p^ to p 2 ', and e is the 
initial (or in situ) void ratio. An approximate correlation 
between CR and natural water content in clays is given 
by the following: 

CR = 0.006 (w-12)(3-5) 

d. Coefficient of volume compressibility. The 
relationship between deformation (or strain) and stress 
for one-dimensional compression is expressed by the 
coefficient of volume compressibility, m , which is 
defined as 



m„ 



Ae 
Ap 

m v = 



^£- 



Ap (1 + e ) 

0.434 a 

(1 + e )p' 



-a^ 



1 +e 



(3-6) 





1 1 1 | ! 1 
— 2 


| 1 1 — 1 | 1 1 j 1 1 1 | 1 1 

* 6 8 SENSITIVITY AT 
\ \ \ LIQUID LIMIT 


1 1 1 | 1 1 


- 1.2 

a. 




\ \ 


V 








\ \ 


\ 






j 
j 


- 


\\ 


\\ 




- 


j 
a. o.e 

3 

■ 

J 












APPROXIMATE MINIM 
OF CONSOLIDATION 
VS. LIQUIDITY INDE> 


UM VALUE 
PRESSURE 






— 


5 o- 4 
□ 

z 














LIQUIDITY 






SENSf 
APPRl 
AT PL 


1VITY ^^O^* 

CACHING I >0 
ASTIC LIMIT — "^^ 












^^ 












^^ 


-0.4 


1 1 1 1 1 1 


1 1 


ill 


1 1 1 1 1 I 


1 1 ill 1 ■ 



O.OI 



0.1 1 10 

EFFECTIVE CONSOLIDATION PRESSURE. TSF 



100 



(NAVFAC DM-7) 

Figure 3-8. Approximate relation between liquidity index and effective overburden pressure, as a function of the sensitivity 

of the soil 



3-9 



where 
Ae 
Ap 

a v 
P' 



change in vertical strain 

P2 - p; = corresponding change in 

effective vertical stress 

Ae/Ap = coefficient of 

compressibility 

average of initial and final 

effective vertical stress 



The unit s of mv ar e the reciprocal of constrained 
modulus. I Table 3^ gives typical values of mv for 
several granular soils during virgin loading. 

e. Expansion and recompression. If 

overburden pressure is decreased, soil undergoes 
volumetric expansion (swell), as shown i n 1 figure 3-71 
The semilogarithmic, straight-line (this may have to be 
approximated) slope of the swelling curve is expressed 
by the swelling index, C s , as 



C s = 



Ae 
log q£ 
Pi' 



TM 5-818-1 / AFM 88-3, Chap. 7 

(3-8) 



where Ae is the change in void ratio (strictly a sign 
applies to C c , C s ., C r , and m v ; however, judgment is 
usually used in lieu of signs). The swelling index is 
generally from one-fifth to one-tenth the compression 
index. Appro ximate values of C s may be obtained from 
I figure 3-1071 The slope of the recompression curve is 
expressed by the recompression index, C r , as follows: 



C r = 



Ae 
logpT 
Pi' 



(3-9) 



The value of C r is equal to or slightly smaller than C s . 
High values of C r /C s are associated with 
overconsolidated clays containing swelling clay minerals. 



is 



«J) 



Si 



is 



2 m 

IS 



10 



05 

















\u 


- 120 
= 80 












Highly 

colloidal 

clays 










\tf: 


80 
50 








Co 
cla 


loidal \ 










i 


v ( LL - 50\ 
\1 PI - 25 \ 

>v Clay* \ 










■N 1 \ 

J LL - 30 
xj PI - 12 v. 
^^ 1 Silty\^ 


\\ 








S 


^lays 
















""""^-^S 


. 



0.001 0.01 0.1 1 10 100 1000 

EFFECTIVE OVERBURDEN PRESSURE, TSF 

I » 1 l I I 1 1 . 1 I I 

IS 10 30 150 300 1000 3000 10.000 

APPROXIMATE DEPTH, FT 

(Courtesy of T. 1/1/. Lambe and R. V. Whitman, 
Soils Mechanics, 1969, p 320. Reprinted by 
permission of John Wiley & Sons, Inc., New York.) 

Figure 3-9. Approximate relation between void ratio and effective overburden pressure for clay sediments, as a function of 

the Atterberg limits. 



3-10 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 3-2. Estimating Degree of Preconsolidation 



Method 



Remarks 



Surface topography 



Geological evidence 



Soil below alluvial valley filling should generally have a preconsolidation stress at 
least corresponding to elevation of abutments. In wide river valleys with 
terraces at several elevations, an elevation corresponding to previous surface 
elevation in the river valley may be several miles distant 

Ask geologist for estimate of maximum preconsolidation stress. Erosion may 
have removed hundreds of feet of material even in abutment area 



Water content 



If natural water content is near PL or below it, anticipate high preconsolidation 
stress. A high natural water content is not, itself, a suitable indicator of 
absence of overconsolidation 



Standard penetration resistance 



If blow counts are high, anticipate high preconsolidation stress. From blow 
counts, estimate undrained shear strength, s u , in tons per square foot as 
approximately 1/15 of blow count. If estimated value is substantially more 



than corresponds to a 
preconsolidation stresses 



s u /p ratio of about 0.25, anticipate high 



Undisturbed sampling 



Laboratory shear strengths 



Compression index from consoli- 
dations tests 



Liquidity index and sensitivity 



If soil was too hard to sample with piston sampler, and a Denison or similar 
sampler was required, suspect high preconsolidation stress 

If higher than those corresponding to a s u /p ratio of about 0.3, anticipate high 
preconsolidation stress 

If compression index appears low for Atterberg limits of soil, suspect that test 
loads were not carried high enough to determine virgin compression curve 
and correct preconsolidation stress. Expected values for compression index 
can be es timated from correlations with water content and Atterberg limits 
Ktable 3-3l 

Estimate preconsolidation stress fronFTfigures 3-8|and 3-9|(p c values may be low) 



U. S. Army Corps of Engineers 



3-11 



f. Coefficient of consolidation. The soil 
properties that control the drainage rate of pore water 
are combined into the coefficient of consolidation, C v , 
defined as follows: 



where 



TM 5-818-1 / AFM 88-3, Chap. 7 



k = coefficient of permeability in a vertical 



direction 



C v = 
alternatively, 



k(1 + e ) 

y w a v 



y w m v 



(3-10) 



c v= 



TH< 



(3-11) 
Table 3-3. Compression Index Correlations 



e = initial void ratio 

y y = unit weight of water 

a v = Ae/Ap = coefficient of compressibility, 

vertical deformation 
m v = coefficie nt of volume compressibility 
T = Time fac tor (para 5-I 5) that depends on 

percent consolidation and assumed 

pore pres- 



Clays 



C c = 0.012 w , wn in percent 
C c = 0.01 (LL-13) 

Sand, uniform 

C c = 0.03 , loose to Cc = 0.06, dense 

Silt, uniform 
C c = 0.20 



U. S. Army Corps of Engineers 

Table 3-4. Value of Coefficient of Compressibility (m v ) for Several Granular Soils During Virgin Loading 



Soil 

Uniform gravel 

1 < D < 5 mm 
Well-graded sand 

0.02 < D < 1 mm 
Uniform fine sand 

0.07 < D<0.3mm 
Uniform silt 

0.02 < D < 0.07 mm 



Relative 
Density, D R 


100 


100 


100 

100 





m v x 10 - 4 
Effective F 


per psi 
'ressure 




9 to 1 4 | 


psi 






28 to 74 psi 


2.3 










1.1 


0.6 










0.4 


5.0 










2.7 


1.3 










0.6 


4.8 










2.0 


1.4 










0.6 


25.0 










4.0 


2.0 










0.9 



(Courtesy of T W. Lambe and R. V. Whitman, Soils 
Mechanics, 1969, p 155. Reprinted by permission of 
John Wiley & Sons, Inc., New York.) 



3-12 



TM 5-818-1 / AFM 88-3, Chap. 7 



0.30 



0.24 



0.20 



O 

X 

Id 
D 

Z 

ID 

Z 

J 
J 
LU 
S 
tf) 



0.15 



0.10 



0.05 



C AT SHRINKAGE LIMIT 



1 1 

CAT PLASTIC LIMIT 



INCREASE OF C s AS REBOUND 
VOID RATIO DECREASES FROM 
LL TO PL FOR SPECIFIC MATERIAL 




0.5 1.0 1.5 2.0 2.5 3.0 

VOID RATIO FROM WHICH REBOUND OCCURS, e. 



3,5 



(NAVFAC DM-7) 



Figure 3-10. Approximate correlations for swelling index of silts and clays. 

3-13 



TM 5-818-1 / AFM 88-3, Chap. 7 



u 

Id 
w 

5 
U 



> 

u 

z 

o 

r- 
< 
a 

j 
o 
</> 
z 
o 
u 

u. 
o 

t- 
z 

Ul 

o 
li. 

1L 

u 
o 
u 



4-10 




- 2 



1 

0.7 
0.5 

0.3 

0.2 

> 
< 
Q 
\ 

CM 
0.* t- 

u. 

0.07 > 
U 

- 0.05 

0.03 
0.02 

0.01 



0.007 



0.005 



80 100 
LIQUID LIMIT (LL) 



120 



140 



160 



Figure 3-11. Correlations between coefficient of consolidation and liquid limit. 



3-14 



TM 5-818-1 / AFM 88-3, Chap. 7 



sure distribution in soil caused by load 
length of longest drainage path (lab or field) 
time at which the time factor is T for the 
degree of consolidation that has occurred 
(generally, use t 50 for T = 0.197 and 50 
percent consolidation) 



Correlation between C v and LL are shown in | figure 3-111 
for undisturbed and remolded soil. 

g. Coefficient of secondary compression. The 
coefficient of secondary compression, C^, is strain e z = 
AH/H , which occurs during one log cycle of time 
following completion of primary consolidation Kfiq 3-71 . 
The coefficient of secondary compression is computed 
as 

AH 

C„ = Ae z , H, (3-12) 



^gT 



to 



log_L 

to 



where t p is time to complete primary consolidation, and 
H f is total thickness of compressible soil at time t p . Soils 
with high compressibilities as determined by the 
compression index of virgin compression ratio will 
generally also have high values of C.,. Highly sensitive 
clays and soils with high organic contents usually exhibit 
high rates of secondary compression. Overconsolidation 
can markedly decrease secondary compression. 
Depending on the degree of overconsolidation, the value 
of C„ is typically about one-half to one-third as large for 
pressures below the preconsolidation pressure as it is for 
the pressures above the preconsolidation pressures. For 
many soils, the value of C x approximately equals 
0.0001 5w, with w in percent. 

h. Effects of remolding. Remolding or 
disturbance has the following effects relative to 
undisturbed soil: 

(1) e-log p curve. Disturbance lowers the 
void ratio reached under applied stresses in the vicinity 
of the preconsolidation stress and reduces the distinct 
break in the curve at the preconsolidation pressure (fig 3- 
7). At stresses well above the preconsolidation stress, 
the e-log p curve approaches closely that for good 
undisturbed samples. 

(2) Preconsolidation stress. Disturbance 
lowers the apparent preconsolidation stress. 

(3) Virgin compression. Disturbance 
lowers the value of the compression index, but the effect 
may not be severe. 

(4) Swelling and recompression. 
Disturbance increases the swelling and recompression 
indices. 

(5) Coefficient of consolidation. 
Disturbance decreases the coefficient of con solidation 
for both virgin compression and recompression Kfiq 3-1 1 ) 
in the vicinity of initial overburden and preconsolidation 
stresses. For good undisturbed samples, the value of 



C v decreases abruptly at the preconsolidation pressure. 

(6) Coefficient of secondary compression. 
Disturbance decreases the coefficient of secondary 
compression in the range of virgin compression. 
3-6. Swelling, shrinkage, and collapsibility. 

a. The swelling potential is an index property 
and equals the percent swell of a laterally confined soil 
sample that has soaked under a surcharge of 1 pound 
per square inch after being compacted to the maximum 
density at optimum water content according to the 
standard compaction test method. Correlation between 
swelling potential and PI for natural soils compacted at 
optimum water conte nt to standard maximum density is 
shown in lfiqure 3-12J 

b. The amount of swelling and shrinkage 
depends on the initial water content. If the soil is wetter 
than the shrinkage limit (SL), the maximum possible 
shrinkage will be related to the difference between the 
actual water content and the SL. Similarly, little swell 
will occur after-the water content has reached some 
value above the plastic limit. 



too 




3-15 



30 40 

Plasticity Index 

(Courtesy of H. B. Seed, J. Woodward. J and 
Lundgren, R., "Predication of Swelling Potential for 
Swelling Clay, " Journal, Soil Mechanics and 
Foundations Division , Vol 88, No. SM3, Part I. 1962, 
pp 53-87. Reprinted by permission of American 
Society of Civil Engineers, New York.) 

Figure 3-12. Predicted relationship between swelling 
potential and plasticity index for compacted soils. 



TM 5-818-1 / AFM 88-3, Chap. 7 



c. Collapsible soils are unsaturated soils that 
undergo large decreases in volume upon wetting with or 
without additional loading. An estimate of collapsibility 
(decrease in volume from change in moistu re available) 
and expansion of a soil may be made from l figure 3-131 
based on in situ dry density and LL. 

3-7. Shear strength of soils. 

a. Undrained and effective strengths. The 
shear strength of soils is largely a function of the 
effective normal stress on the shear plane, which equals 
the total normal force less the pore water pressure. The 
shear strength, s, can be expressed in terms of the total 
normal pressure, a, or the effective normal pressure, a', 
by parameters determined from laboratory tests or, 
occasionally, estimated from correlations with index 
properties. The shear test apparatus is shown ir l fiqurel 
3-14. The equations for shear strength are as follows: 

s = c + tan (j) (total shear strength parameters) 
s' = c' + (a - u) tan ty (effective shear 
strength parameters) 



The total (undrained) shear strength parameters, c and 
ty, are designated as cohesion and angle of internal 
friction, respectively. Undrained shear strengths apply 
where there is no change in the volume of pore water 
(i.e., no consolidation) and are measured in the 
laboratory by shearing without permitting drainage. For 
saturated soils, (j> = 0, and the undrained shear strength, 
c, is designated as s u . The effective stress parameters, 
c' and (j)', are used for determining the shear strength 
provided pore pressures, u, are known. Pore pressure 
changes are caused by a change in either normal or 
shear stress and may be either positive or negative. 
Pore pressures are determined from piezometer 
observations during and after construction or, for design 
purposes, estimated on the basis of experience and 
behavior of samples subjected to shear tests. Effective 
stress parameters are computed from laboratory tests in 
which pore pressures induced during shear are 
measured or by applying the shearing load sufficiently 
slow to result in fully drained conditions within the test 
specimen. 



so 



U-70 

a. 
y 

h 

at ao 
z 
ui 
a 

> 

£ 90 
□ 

J 

< 

E 

^100 

< 

z 



I I 


I 


I 


i \^ 




COLLAPSE 






EXPANSION 








' 


VERY HIGH 




- 




HIGH 




- 


// 


I 

(medium 






- 




I 








t, //'' LOW 

// i i 


I 

I 

I 

I, 


. 


, 


- 



to 



40 SO 

LIQUID LIMIT 



60 



70 



80 



90 



(Courtesy of J. K. Mitchell and W. S. Gardner, "In Situ 
Measurement of Volume Change Characteristics: 
Geotechnical Engineering Division Specialty Conference on 
In Situ Measurement of Soil Properties , 1975. North 
Carolina State University, Raleigh, N. C. Reprinted by 
permission of American Society of Civil Engineers, New 
York.) 

Figure 3-13. Guide to collapsibility, compressibility, and expansion based on in situ dry density and liquid limit. 



3-16 



TM 5-818-1 / AFM 88-3, Chap. 7 









¥ 









^ - »ueS>."J*NQ 



"fr=HM 



3 



SCHEMATIC DIAGRAM OF TRIAX1AL APPARATUS 




A SAMPLE 

6 LOWER FRAME 

C UPPER FRAME 

D POROUS BRONZE PLATE 

E ORAINASE CHANNELS 

F PISTON 

G WATER RESERVOIR 

H BASE 

I PRINCIPAL PLANEOF FAILURE 

K SECONDARY PLANES OF FAILURE 

P NORMAL FORCE 

S SHEARINQ FORCE 

SCHEMATIC DIAGRAM OF DIRECT SHEAR APPARATUS 




TYPICAL DESIGH 
SMEAR STRENGTHS 



IANGE OF SHEAR STRENGTHS 
FROM LABORATORY TESTS 



a. SHEAR TEST APPARATUS 



NORM AL STRESS 

C. SHEARING RESISTANCE OF SANDS 



1.5 

c_ i. a 
! 

0.5 





10 1 S 20 28 30 39 40 49 

0*>- DEGREES 
EFFEC Tl V E 
CONSOLIDATION 
PRESSURE, p_— m \ 




UNCON FIN ED 
ST REN GTM 



R API D TRI A XI AL 
STRENGTH 



QUICK TRI AXI ALl 
STRENGTH 



SLOW TRI A XI AL STRENGTH 

t>. TYPICAL FAILURE ENVELOPES FOR Q, R, AND S TESTS 




RESIDUAL 



DISPLACEMENT, b 



d. SHEARING RESISTANCE OF OVER-CONSOLIDATED (OC> AND 
NORMALLY-CONSOLIDATED (NO CLAYS 



Figure 3-14. Shear test apparatus and shearing resistance of soiis. 

3-17 



TM 5-818-1-1 / AFM 88-3, Chap. 7 



b. Undrained shear strength-cohesive soils. 
Approximate undrained shear strengths of fine-grained 
cohesive soils can be rapidly determined on undisturbed 
samples and occasionally on reasonably intact samples 
from drive sampling, using simple devices such as the 
pocket penetrometer, laboratory vane shear device, or 
the miniature vane shear device (Torvane). To establish 
the reliability of these tests, it is desirable to correlate 
them with unconfined compression tests. Unconfined 
compression tests are widely used because they are 
somewhat simpler than Q triaxial compression tests, but 
test results may scatter broadly. A more desirable test is 
a single Q triaxial compression test with the chamber 
pressure equal to the total in situ stress. Unconfined 
compression tests are appropriate primarily for testing 
saturated clays that are not jointed or slickensided. The 



Q triaxial compression test is commonly performed on 
foundation clays since the in situ undrained shear 
strength generally controls the allowable bearing 
capacity. Sufficient unconfined compression and/or Q 
tests should be performed to establish a detailed profile 
of undrained shear strength with depth. Undrained 
strengths may also be estimated from the standard 
penetration test, cone penet rometer sou ndings, and field 
vane tests, as discussed i h chapter"!! . For important 
structures, the effects of loading or unloading on the 
undrained shear strength should be determined by R 
(consolidated-undrained) triaxial compression tests on 
representative samples of each stratum. 

c. Strength parameters, cohesive soils. The 
undrained shear strength of saturated clays can be 
expressed as 



4.0 



3.5 



3.0 



2.5 

X 

W 
Q 

? 2.0 

>- 
I- 



D 1.5 
O 



1.0 



0.9 



-0.5 













1 * *\ 










\ * * 


V S APPRO, 
JT LIMITS < 


KIMATE 
■)F DATA 






V 


\ 




















x '"• 














^w 






















• 



0.0001 



0.001 



0.01 



0.1 



1 



10 



REMOLDED SHEAR STRENGTH, KG/CM' 



(Courtesy of W. N. Houston and J. K. Mitchell, "Property 
Interrelationships in Sensitive Clays," Journal, Soil 
Mechanics and Foundations Division , Vol 95, No. SM4. 
1969, pp 1037-1062. Reprinted by permission of 
American Society of Civil Engineers. New York.) 

Figure 3-15. Remoltded shear strength versus liquidity index relationships for different clays. 



3-18 



TM 5-818-1 / AFM 88-3, Chap. 7 



Vo/OC 4 

Wnc 





I I I I 

RESULTS OF DIRECT SIMPLE SHEAR 
TESTS ON 5 COHESIVE SOILS WITH 


1 1 1 1 


1 




( — ) = 0.17 - 0.29 
\P /NC 




^^y^ — 


_ 


PI = 20 - 75 


^^<^ r V > >. 


^^y^ — 


- 


y!^^^-' 


L»<^V '•\Ly***^^ AVERAGE 

^APPROXIMATE 
LIMITS OF DATA 


— 




y^^y^"^ 






- 


1 1 1 1 


1 1 1 1 


1 



5 b 7 

OVERCONSOLIDATION RATIO 



t1 



(Courtesy of C C. Ladd and R. Foott, "New Design Procedure for Stability of Soft Clays, " Journal, 
Geotechnical Engineering Division , Vol 100, No. GT7, 1974, pp 763-786. Reprinted by permission 

of American Society of Civil Engineers, New York.) 

Figure 3-16. Normalized variation of sjpo ratio for overconsolidated clay. 



3-19 



TM 5-818-1 / AFM 88-3, Chap. 7 



Su 



2 



s u = C u (j) u = 

(01 - 03) = 



(3-11) 

*- (3-12) 
2 



and is essentially independent of total normal stress. 
The undrained cohesion intercept of the Mohr-Cou-lomb 
failure envelope is c u . 

(1) The undrained shear strength, s, , of 
normally consolidated cohesive soils is proportional to 
the effective overburden pressure, po. An approximate 
correlation is as follows: 



s u =0.11 +0.0037 PI 
T^~ 



(3-13) 



(2) A correlation between the remolded, 
undrained shear streng th of clays and the liquidity index 
is shown in lfigure 3-151 

(3) A correlation between the normalized 
s u /p ratio of overconsolidated soils and the 
overconsolidation ratio (OCR) is presented i n 1 figure 3-T6] 
The value of p in (s u /p )OC is the effective present 
overburden pressure. Values of (s u /p ) may be 
estimated from this figure when (s u /p )NC and the OCR 
are known (NC signifies normally consolidated soils). 

d. Sensitivity, cohesive soils. The sensitivity 
of a clay soil, St, is defined as follows: 



Terms descriptive of sensitivity are listed in I table S^l 
Generalized relationships among sensitivity, liquidity 
index, and e ffective overburden pressure are shown in 
I figure 3-171 The preconsolidation pressure, rather than 
the effective overburden pressure, should be used for 
overconsolidated soils when entering this figure. 
Cementation and aging c ause higher values of sensitivity 
than given in lfigure 3-TTl 

e. Effectiv e strength p arameters, cohesive 
soils. As indicated in lfigure 3-14,1 the peak and residual 
strengths may be shown as failure and postfailure 
envelopes. Values of the peak drained friction angle for 
normally co nsolidated clays may be estimated from 

I figure 3-i"8~l After reaching the peak shear strength, 
overconsolidated clays strain-soften to a residual value 
of strength corresponding to the resistance to sliding on 
an established shear plane. Large displacements are 
necessary to achieve this minimum ultimate strength 
requiring an annular shear apparatus or multiple 
reversals in the direct shear box. Typical val ues of 
residual angles of friction are shown in lfigure 3 : W] 

f. Shear strength, cohesionless soils. 

(1) In sandy soils, the cohesion is 
negligible. Because of the relatively high permeability of 
sands, the angle of internal friction is usually based 
solely on drained tests. The angle of internal friction of 
sand is primarily affected by the density of the sand and 
normally varies within the limits of about 28 to 46 de- 



S t = 

(at 



Undisturbed compressive strength 
Remolded compressive strength 
same water 



content) 



Table 3-5. Sensitivity of Clays 



Sensitivity 
0-1 



Descriptive Term 
Insensitive 



1-2 

2-4 

4-8 

8-16 

>16 



Low sensitivity 
Medium sensitivity 
Sensitive 
Extra sensitive 
Quick 



U. S. Army Corps of Engineers 



3-20 



TM 5-818-1 / AFM 88-3, Chap. 7 



orees lffig 3-3i Approximate values of (j> are given as 
follows: 

4>= 30 + 0.15 DR for soils with less than 5 

percent fines(3-14) 
4> = 25 + 0.15 DR for soils with more than 5 
percent fines(3-15) 

Values of <j) = 25 degrees for loose sands and ({> = 35 
degrees for dense sands are conservative for most 
cases of static loading. If higher values are used, they 
should be justified by results from consolidated- drained 
triaxial tests. 

(2) Silt tends to be dilative or contractive 
depending upon the consolidation stresses applied. 



Thus, the back-pressure saturated, consolidated- 
undrained triaxial test with pore pressure measurements 
is used. If the silt is dilative, the strength is determined 
from the consolidated-drained shear test. The strength 
determined from the consolidated-undrained test is used 
if the silt is contractive. Typical values of the angle of 
internal friction from consolidated-drained tests 
commonly range from 27 to 30 degrees for silt and silty 
sands and from 30 to 35 degrees for loose and dense 
conditions. The consolidated-undrained tests yield 15 to 
25 degrees. The shear strength used for design should 
be that obtained from the consolidated-drained tests. 



2.0 



X 

tu 

Q 

Z 

> 

I- 



O 




1 10 

EFFECTIVE STRESS, TSF 



100 



(Courtesy of W. N. Houston and J. K. Mitchell, "Property 
Interrelationships in Sensitive Clays, " Journal, Soil 
Mechan- ics and Foundations Division , Vol 95, No. SM4, 
1969, pp 1037-1062. Reprinted by permission of 
American Society of Civil Engineers, New York.) 

Figure 3- 1 7. General relationship between sensitivity, liquidity index, and effective overburden pressure. 



3-21 



TM 5-818-1 / AFM 88-3, Chap. 7 



50 



« «0 



o 



30 



Z 20 

< 



y io 

<r 



"■T T 
/ 


i ■■ t rr 




1 i — 1 i i i i 

Average ! 1 Std. Dev. 


f ~ 


<n 

— ~s^~ 


T—-S. 


DO * 
Sffitev o 








a 


Q 


^**« 
















t>'at (af/o5W« 


o 
o 

A 


1 ' 


till. 








i 1 i i i i 



5 10 20 SO 

PLASTICITT INDEX, PI (%) 



too 



U. S. Army Corps of Engineers 

Figure 3- 18. Empirical correlation between friction angle 

and plasticity index from triaxial compression tests on 

normally consolidated undisturbed clays. 

3-8. Elastic properties (E, n). The elastic modulus 
and Poisson's ratio are often used in connection with the 
elasticity theory for estimating subsoil deformations. 
Both of these elastic properties vary nonlinearly with 
confining pressure and shear stress. Typical values 
given below refer to moderate confining pressures and 
shear stresses corresponding to a factor of safety of 2 or 
more. 

a. In practical problems, stresses before 
loading are generally anisotropic. It is generally 
considered that the modulus of elasticity is proportional 
to the square root of the average initial principal stress, 
which may usually be taken as 



°v 



1 +2K 
3 



1/2 



(3-16) 



where K is the coefficient of at-rest earth pressure fparal 
13-10)1 and 5 V ' is the effective vertical stress. This 
proportionality holds for 0.5 < K < 2, when working 
stresses are less than one-half the peak strength. 

b. The undrained modulus for normally 
consolidated clays may be related to the undrained shear 
strength, su, by the expression 



E_ 

Su 



= 250 to 500 



(3-17) 



where s u is determined from Q tests or field vane shear 
tests. The undrain ed modulus may also be estimated 
from Ifiqure 3-20. I Field moduli may be double these 
values. 

c. Poisson's ratio varies with strain and may 



be as low as 0.1 to 0.2 at small strains, or more than 0.5. 
3-9. Modulus of subgrade reaction. 

a. The modulus of subgrade reaction, k s , is 
the ratio of load intensity to subgrade deformation, or: 
k s = q (3-18) 



where 

q = intensity of soil pressure, pounds or 

kips per square foot 
A = corresponding average settlement, 
feet 
b. Values of k. may be obtained from general 
order of decreasing accuracy: 

(1 ) Plate or pile load test (chaps 4 and 1 2). 

(2) Empirical equations (additional discussion 
in chap 10). 

(3) Tabulated values fTtabie"^ . 

3-10. Coefficient of at-rest earth pressure. The 

state of effective lateral stress in situ under at-rest 
conditions can be expressed through the coefficient of 
earth pressure at rest and the existing vertical 
overburden pressure. This ratio is termed Ko and given 
by the following: 



K n 



o h _ 



(3-19) 



The coefficient of at-rest earth pressure applies for a 
condition of no lateral strain. Estimate values of K as 
follows: 

Normally consolidated soil 
Sand: 

K = 1 - sin f (3-20) 

Clay: 

K = 0.95 - sin f (3-21) 

I Figure 3-2T| may be used for estimates of K for both 
normally consolidated and overconsolidated soils in 
terms of PI. For overconsolidated soils, this figure 
applies mainly for unloading conditions, and reloading 
may cause a large drop in K values. For soils that 
display high overconsolidation ratios as a result of 
desiccati on, K will b e overestimated by the relationship 
shown in lfigure 3-21 J 

3-11. Properties of intact rock. The modulus ratio 
and uniaxial compressiv e strength of various intact rocks 
are shown in Itable 2 -7J 

3-12. Properties of typical shales. Behavioral 
characteristics of shales are summarized in l table j£Z] 
and physical properties of various shales, in l table 3-"8l 
Analyses of observed in situ behavior provide the most 
reliable means for assessing and predicting the behavior 
of shales. 



3-22 



40 



30 — 



111 

J 
O 

T 
< 

Z 

O 

H 20 

O 

a 
k. 

j 
< 

D 
O 
m 
£ '0 



TM 5-818-1 AFM 88-3, Chap. 7 



I 


I 


I 




i i i 

CLAY SHALES 

AVERAGE AND RANGE FOR NATURAL SOILS 


- 


WA 




- 




^%/^^^ 




- 


I 




I 


i 


O '^^~ C -' ' ' t / / &. 

I I I 





30 



SO 



PLASTICITY INDEX 



60 



70 



Figure 3-19. Relation between residual friction angle and plasticity index. 



80 



3-23 



TM 5-818-1 / AFM 88-3, Chap. 7 



1600 



1400 



1200 



1000 



* 800 



600 



400 



200 



1 1 I I I I 



E u - UNDRAINED MODULUS OF CLAY 
K - FACTOR FROM CHART _ 

S„ - UNDRAINED SHEAR STRENGTH 
OF CLAY 




1 1.5 2 3 4 5 6 7 8 9 10 

OVERCONSOLIDATION RATIO 

U. S. Army Corps of Engineers 

Figure 3-20. Chart for estimating undrained modulus of clay. 
Table 3-6. Values of Modulus of Subgrade Reaction (k s ) for Footings as a Guide to Order of Magnitude 



Soil Type 



Loose sand 

Medium sand 

Dense sand 

Clayey sand (medium) 

Silty sand (medium) 

Clayey soil 

q u < 4ksf 

4<q u < 8 

8<q u 



Range of k s , kef 
30-100 
60-500 
400-800 
200-500 
150-300 

75-150 

150-300 

>300 



a Local values may be higher or lower. 
U. S. Army Corps of Engineers 



3-24 



TM 5-818-1 / AFM 88-3, Chap. 7 



3.0 
o 



5 2.0 
i/> 

uu 

a. 

I 1.5 

>— 

< 

O 10 

»— 

z 

UJ 

£ 05 

UJ 

o 





► , 


• 














( ^*^ • 




















*s^ Ov«rcon»otic 


arion Ratio -OCR. 










• 


s^32 








'f 


















*^"^ * 






^J6 








n 








6 
























j> 


r • 






4 














— » 


c 








__J____ 














j 








_ 1 














-A— _ — 

















10 20 30 40 50 

PLASTICITY INDEX. PI 



60 



70 



80 



(Reproduced by permission of the National Research 
Council of Canadafrom the Canadian Geotechnical 
Journal, Volume 2, pp 1-15, 1965.) 

Figure 3-21. Coefficient of earthpressure at rest (K ) as a function of overconsolidated ratio and pasticity index. 



3-25 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 3- 7. An Engineering Evaluation of Shales 



Physical Properties 


Probable In-situ Behavior 


Laboratory tests 
and in-sltu 


Average range of values 


High 
pore 
Pres- 
sure 

<<) 


Low 
bearing 
capacity 

(5) 


Tendency 

to 
rebound 

<6) 


Slope 
stability 
problems 

(7) 


Rapid 
slaking 

(8) 


Rapid 
erosion 

(9) 


Tunnel 
support 
prob- 
lems 

(10) 


observations 
(1) 


Unfavorable 
(2) 


Favorable 
(3) 


Compressive 

strength, In 

pounds per square 

Inch 


50 to 300 




V 


V 














300-5000 
















Modulus of 
elasticity, in 


20,000 to 
200,000 






V 










V 


pounds per 
square inch 




200,000 to 
2x 10* 
















Cohesive strength, 


5 to 100 








_.v 


. _v__ 






V 


In pounds per 
square inch 




100 to 
>1500 
















Angle of internal 


10 to 20 








__v 


_ V . 






V 


friction, In degrees 




20 to 65 
















Dry density, In 


70 to 110 




-tf- 










Vw 




pounds per cubic 
foot 




110 to 160 
















Potential swell. 


3 to 15 








V 


V 




V 


V 


in percentage 




1 to 3 
















Natural Moisture 


20 to 35 




_tf_ 






V 








content. In 
percentage 




5-15 
















Coefficient of 
permeability, in 


10"» to 
io- ,0 




V 






y/ 


V 






centimeters per 
second 




>10" 8 
















Predominant clay 
minerals 


Montmoril- 

lonite or 

Illlte 




tf_ 






V 










Kaollnite & 
Chlorite 
















Activity ratio 
_ Plasticity index 
Clay content 


0.75 to>2.0 










V 










0.35 to 0.75 
















Wetting and 


Reduces to 
grain sizes 












V 


V 




drying cycles 




Reduces to 
Flakes 
















Spacing of 


Closely 
Spaced 






V 




V 




V(?i 


. V . 


rock defects 




Widely 
Spaced 
















" Orientation of 


Adversely 
Oriented 






V 




V 






V 


rock defects 




Favorably 
Oriented 
















State of stress 


>Existing 
Over- 
burden 
Load 








V 


V 






V 






ar Over- 
burden 
Load 

















(Courtesy of L. B. Underwood, "Classification and Identification of 
Shales. " Journal, Soil Mechanics and Foundations Division, Vol 93, 
No. SM6, 1967, pp 97-116. Reprinted by permission of American 

Society of Civil Engineers, New York.) 



3-26 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 3-8. Physical Properties of Various Shales 



Name of Shale 
Formation, Age, Locality 

(1) 


Compres- 
sive 
strength, 
in pounds 
per square 
inch 
(2) 


Modulus of 

Elasticity, 

in pounds per 

square inch 

(3) 


Cohesion, 

in pounds 

per square 

inch 

(4) 


Angle of 

Internal 

Friction, 

in degrees 

(5) 


Dry 
Density, In 
pounds per 
cubic foot 

(6) 


Potential 

Swell In 

percentage 

(7) 


Natural 
Moisture In 
percentage 

(8) 


Predominant 

Clay 

Minerals 

(9) 


Activity 
Ratio 

(10) 


Bearpaw, Cret. , Canada 
Weathered - 


7 to 84 


7500 


3 to6 


6 to 20 


85-95 


0.5%-2% 


29 to 36 


nine, 

Montmoril- 
lonlte 


0.30 to 


Unweathered- 


154-406 


18,000 


22 


30 


95-108 


5%-20% 


19 to 27 


Mixed-layer 


>1.5 


Pierre, Cret., 
So. Dakota 


70-1400 


20,000-140,000 


2 to 30 


8-25 


95-110 


3% to 5% 


18 to 27 


do 


0.3 to>2 


Ft. Union, Tert., 


70-1050 


11,200-56,000 


10 


20 


95-115 


2%<?) 


16-24 


niite 




No. Dakota 




Pepper, Cret., Texas 


28-70 




2-6 


7-14 


110 




20% 


mite Mont. 


1.2 


56-154 




1-8 


19-28 


119 




17% 


Kaolinite 
Illlte, Kao- 


1.0 


Trinity, Cret., Texas 


30-170 


2,400-33,000 


0-7 


26 


115-133 




11-17% 


linite, Mont. 






250-550 


6,000-20,000 
1,000,000 


1 5-25 


8-30 


112-118 




15%- 18% 






Composite 


" Silty Clayey 
Carbonaceous 

Clay Bonded 

Clayey 
Ferruginous 


210 
4165 

2084 

1661 
3674 


56 
1562 

931 

488 
1600 


23 
16 

7 

29 

9 


138 




9.1 






Cyclotnem 
of Peim- 
sylvanlan 
















(Eastern 
Ohio and 










Penn.) 


486,500 










Niobrara, 

Calc. Sh.,Cret., Colo. 














Illlte, Beld. 




Mowry, Cret., Colo. 
















Kaolinite, 
Chlorite 

mite 




Graneros, Cret., Colo. 
















Kaol.,111., 
Mix-layer 




Morrison, Jura, Colo. 
















Mont., Illlte 




Laramie, Cret. , Colo. 
















niite, Beld. 




Mauv, Calc. Shale, 


5220 


2.3 x 10* 


1,160 
3,390 


64 


164 




2% 






Cam., Utah 
Quartzose Sh., 


17,770 
bs. per Inch 


2.3 x 10* 


45 


165 




4% 






Cam., Utah 

1 Ton/sq. ft. » approx. 14 

















(Courtesy of the American Society of Civil Engineers) 



3-27 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 4 



FIELD EXPLORATIONS 



4-1. Investigational programs. Field 

investigations can be divided into two major phases, a 
surface examination and a subsurface exploration: 
Surface Examination Subsurface Exploration 
Documentary evidence Preliminary 

Field reconnaissance Detailed 

Local experience 

a. Documentary evidence. The logical and 
necessary first step of any field investigation is the 
compilation of all pertinent information on geological and 
soil conditions at and in the vicinity of the site or sites 



under consideration, including previo us excava tions, 
material storage, and buildings. Use l table 4-1 l as a 
guide to sources and types of documentation. 

b. Field reconnaissance. A thorough visual 
examination of the site and the surrounding area by the 
foundation engineer is essential. This activity may be 
combined with a survey of local experience. The field 
reconnaissance should include an examination of the 
following items as appropriate: 

(1) Existing cuts (either natural or man- 
made). Railway and highway cuts, pipeline trenches, and 



Table 4-1. Types and Sources of Documentary Evidence 



Types 

Topographic, soil, drift 

(overburden), and bedrock 

maps 

Surface and subsurface 

mining data, present and 
past 



Aerial photographs, Conti- 
ental U. S. 



Sources 

Local, state, Federal, and uni- 
versity geologic and agricul- 
tural organizations 

U. S. Bureau of Mines and State 
mining groups 



U. S. Government Printing Office 



Descriptions 

These maps provide information on lay 
of land, faulting (tectonics), and 
material types 

Such data help locate subsurface 

shafts and surface pits. The pres- 
ence of cavities in the foundation 
must be known. Current and even old 
workings may represent material 
sources for construction. In addi- 
tion, surface pits near site may pro- 
vide opportunity to observe stratifi- 
cation of foundation and allow taking 
of disturbed samples 

Aerial photos offer a valuable means of 
establishing some insight into the 
nature of foundation soils (2) and 
also expedite familiarization with 
the lay of the land 



Aerial photographs, county 
and state areas 



S. Soil Conservation Service, 
local or district office 



Local experience 



Boring logs, water-well 

records, and construction 
records 



Hydrological and tidal data 



Technical Journals and published 
reports; professional soci- 
eties, universities, and state 
agencies 

State Building Commission, City 
Hall, County Court House, pri- 
vate concerns 



State agencies, river boards, 

U. S. Coast and Geodetic Sur- 
vey, and National Weather 
]Service 



May include considerable boring data, 
test data, and descriptions of 
problems in construction 



Some of these types of information can 
usually be obtained for existing 
public buildings and facilities. 
Private firms may cooperate in pro- 
viding limited data 

Flood history, groundwater levels, and 
tidal data indicate protective meas- 
ures required during and after con- 
struction.Any groundwater informa- 
tion may aid in dewatering facilities 
and safe excavation slopes 



U. S. Army Corps of Engineers 



4-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



walls of river or stream valleys may reveal stratigraphy 
and offer opportunities to obtain general samples for 
basic tests, such as Atterberg limits and grain-size 
analysis for classification. 

(2) Evidence of in situ soil performance. A 
study of landslide scars contributes greatly to the design 
of excavation slopes; it may indicate need for bracing or 
suggest slope maintenance problems because of 
groundwater seepage. Evidence of general or localized 
subsidence suggests compressible subsoils, subsurface 
cavities, or ongoing sink-hole formations as in areas of 
limestone formations or abandoned mine cave-ins. Fault 
scarps or continuous cracks suggest bedrock 
movements or mass soil movements. 

(3) Existing structures. Careful 
observation of damage to existing structures, such as 
cracks in buildings (or poor roof alignment), misaligned 
power lines, pavement conditions, corrosion on pipelines, 
or exposed metal and/or wood at water lines, may 
suggest foundation problems to be encountered or 
avoided. 

(4) Groundwater. The extent of 
construction dewatering may be anticipated from factors 
such as the general water level in streams, spring lines, 
marshy ground, and variations in vegetal growth. The 
effects of lowering the water table during dewatering on 
surrounding structures, as well as potential 
environmental effects, should be appraised in a 
preliminary manner. Drainage problems likely to be 
encountered as a result of topography, confined working 
space, or increased runoff onto adjacent property should 
be noted. 

(5) Availability of construction materials. 
The availability of local construction material and water is 
a major economic factor in foundation type and design. 
Possible borrow areas, quarries and commercial 
material sources, and availability of water should be 
noted. 

(6) Site access. Access to the site for 
drilling and construction equipment should be appraised, 
including the effects of climate during the construction 
season. 

(7) Field investigation records. 
Considering the value and possible complexity of a field 
investigation, a well-kept set of notes is a necessity. A 
camera should be used to supplement notes and to 
enable a better recall and/or information transfer to 
design personnel. 

c. Local experience. Special attention should 
be given to the knowledge of inhabitants of the area. 
Farmers are generally well informed about seasonal 
changes in soil conditions, groundwater, and stream 
flood frequencies. Owners of adjacent properties may 
be able to locate filled areas where old ponds, lakes, or 
wells have been filled, or where foundation of 
demolished structures are buried. 

d. Preliminary subsurface exploration. The 



purpose of preliminary subsurface explorations is to 
obtain approximate soil profiles and representative 
samples from principal strata or to determine bedrock or 
stratigraphic profiles by indirect methods. Auger or 
splitspoon borings are commonly used for obtaining 
representatives samples. Geophysical methods together 
with one to several borings are often used in preliminary 
exploration of sites for large projects, as they are rapid 
and relatively cheap. Procedures for geophysical 
exploration are described in standard textbooks on 
geotechnical engineering. Borings are necessary to 
establish and verify correlations with geophysical data. 
Preliminary reconnaissance explorations furnish data for 
planning detailed and special exploration of sites for 
large and important projects. The preliminary exploration 
may be sufficient for some construction purposes, such 
as excavation or borrow materials. It may be adequate 
also for foundation design of small warehouses, 
residential buildings, and retaining walls located in 
localities where soil properties have been reasonably 
well established as summarized in empirical rules of the 
local building code. 

e. Detailed subsurface explorations. For 
important construction, complex subsurface conditions, 
and cases where preliminary subsurface explorations 
provide insufficient data for design, more detailed 
investigations are necessary. The purpose is to obtain 
detailed geologic profiles, undisturbed samples and 
cores for laboratory testing, or larger and fairly 
continuous representative samples of possible 
construction materials. Test pits and trenches can be 
used to depths of 15 to 25 feet by using front-end 
loaders or backhoes at a cost that may compare 
favorably with other methods, such as auger borings. 
Test pits allow visual inspection of foundation soils; also, 
high-quality undisturbed block samples may be obtained. 
Continuous (2 1/2 - to 5-foot intervals) sampling by 
means of opendrive, piston, or core-boring samplers is 
used for deeper explorations. Penetration, sounding or 
in situ tests, such as vane shear, or pressuremeter tests 
may be conducted depending on sampling difficulty or 
desired information. 
4-2. Soil boring program. 

a. Location and spacing. Borings spaced in a 
rigid pattern often do not disclose unfavorable 
subsurface conditions; therefore, boring locations should 
be selected to define geological units and subsurface 
nonconformities. Borings may have to be spaced at 40 
feet or less when erratic subsurface conditions are 
encountered, in order to delineate lenses, boulders, 
bedrock irregularities, etc. When localized building 
foundation areas are explored, initial borings should be 
located near building corners, but locations should allow 
some final shifting on the site. The number of borings 
should never be less than three and preferably five-one 
at each corner and one at the center, unless subsurface 
conditions are known to be uniform and the 



4-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



foundation area is small. These preliminary borings 
must be supplemented by intermediate borings as 
required by the extent of the area, location of critical 
loaded areas, subsurface conditions, and local practice. 
b. Depth of exploration. The required depth of 
exploration may be only 5 to 10 feet below grade for 
residential construction and lightly loaded warehouses 
and office buildings, provided highly compressible soils 
are known to not occur at greater depths. For important 
or heavily loaded foundations, borings must extend into 
strata of adequate bearing capacity and should penetrate 
all soft or loose deposits even if over- lain by strata of 
stiff or dense soils. The borings should be of sufficient 
depth to establish if groundwater will affect construction, 
cause uplift, or decrease bearing capacity. When 
pumping quantities must be estimated, at least two 
borings should extend to a depth that will define the 
aquifer depth and thickness. Borings may generally be 
stopped when rock is encountered or after a penetration 
of 5 to 20 feet into strata of exceptional stiffness. To 
assure that boulders are not mistaken for bedrock, rock 
coring for 5 to 10 feet is required. When an important 
structure is to be founded on rock, core boring should 
penetrate the rock sufficiently to determine its quality and 
character and the depth and thickness of the weathered 
zone. Rock coring is expensive and slow, and the 
minimum size standard core diameter should be used 
that will provide good cores. NX or larger core sizes may 
be required in some rock strata. Core barrels can 
remove cores in standard 5-, 10-, and 20-foot le ngths 
(actual core may be much fractured, however; see l paral 



1 2-61 Detailed exploration should be carried to a depth 
that encompasses all soil strata likely to be significantly 
affected by structure loading. If the structure is not 
founded on piles, the significant depth is about 1 1/2 to 2 
times the width of the loaded area. An estimate of the 
required depth can be made using the stress influence 
charts in l chapter 5l to find the depth such that 
Aq <=0.1q (4-1) 

where Aq represents an increase in strata stress and q 
is the foundation contact pressure. Note that in the case 
of a pile foundation, stresses are produced in the ground 
to an appreciable depth below the tips of the piles. 
Procedures to obtain Aq apply as for other foundations. 
This depth criterion may not be adequate for complex 
and variable subsurface conditions. 

c. Plugging borings. All borings should be 
carefully plugged with noncontaminating material if- 

(1) Artesian water is present or will be 
when the excavation is made. 

(2) Necessary to avoid pollution of the 
aquifer from surface infiltration, leaching, etc. 

(3) Necessary to preserve a perched 
water table (avoid bottom drainage through borehole). 

(4) Area is adjacent to stream or river 
where flood stage may create artesian pressure through 
the borehole. 

d. Sample requirements. I Table 4-2~| mav be 
used as a g uide for req uired sizes of undisturbed 
samples, and I table 4-3~1 for general samples. The 
sampling program 



Table 4-2. Recommended Undisturbed Sample Diameters 



Test 
Unit weight 
Permeability 
Consolidation 
Triaxial compressiona 
Unconfined compression 
Direct shear 



Minimum 


Sample 


Diameter, 


in 




3.0 








3.0 








5.0 








5.0 








3.0 








5.0 







a Triaxial test specimens are prepared by cutting a short section of 5-in.-diam sample axially into four 
quadrants and trimming each quadrant to the proper size. Three quadrants provide for three tests 
representing the same depth; the fourth quadrant is preserved for a check test. 



U. S. Army Corps of Engineers 



4-3 



TM 5-818-1 / AFM 88-3, Chap. 7 



may depend on drilling equipment available and 
laboratory facilities where tests will be performed. 

(1) Undisturbed samples. Any method of 
taking and removing a sample results in a stress change, 
possible pore water change, and some structure 
alteration because of displacement effects of the 
sampler. Careful attention to details and use of proper 
equipment can reduce disturbance to a tolerable amount. 
Sample disturbance is related to the area ratio A, , 
defined as follows: 



A r = D i-Dr 



X 100 percent (4-2) 



where 



D = outside diameter of sampler tube 

D! = internal diameter of the cutting shoe 

through which the sample passes (commonly the cutting 

edge is swedged to a lesser diameter than the inside 

tube wall thickness to reduce friction) 



The area ratio should be less than 10 percent for 
undisturbed sampling. Undisturbed samples are 
commonly taken by thin-wall seamless steel tubing from 
2 to 3 inches in diameter and lengths from 2 to 4 feet. 
Undisturbed samples for shear, triaxial, and 
consolidation testing are commonly 3 inches in diameter, 
but 5-inch-diameter samples are much preferred. An 
indication of sample quality is the recovery ratio, L r , 
defined as follows: 

Lr = Length of recovered sample (4-3) 
Length sample tube pushed 

A value for L r < 1 indicates that the sample was 
compressed or lost during recovery, and L r > 1 indicates 
that the sample expanded during recovery or the excess 
soil was forced into the sampler. 

(2) Representative samples. Samples can be 
obtained by means of auger or drive-sampling methods. 
Thick-wall, solid, or split-barrel drive samplers can be 
used for all but gravelly soils. Samples taken with a 



Table 4-3. Recommended Minimum Quantity of Material for General Sample Laboratory Testing 



Test 



Minimum Sample Required 
lb (Dry Weight) 3 



Water content 
Atterberg limits 
Shrinkage limits 
Specific gravity 
Grain-size analysis 
Standard compaction 
Permeability 
Direct shear 
4-in.-diam consolidation 
1.4-in.-diam triaxial (4 points) 
2.8-in.-diam triaxial (4 points) 
6-, 12-, or 15-in.-diam 

triaxial (4 Points) 
Vibrated density 



0.5 
0.2 
0.5 
0.2 
0.5 
30.0 
2.0 
2.0 
2.0 
2.0 
8.0 
Discuss with laboratory 

Discuss with laboratory 



Fine grained (all minus No. 4 sieve). For material containing plus No. 4 sieve sizes, the sampling requirements should 
be discussed with the laboratory. In the final analysis, it is the responsibility of the engineer requesting the tests to 
ensure that adequate size samples are obtained. Close coordination with the testing laboratories is essential. 



U. S. Army Corps of Engineers 



4-4 



TM 5-818-1 / AFM 88-3, Chap.7 



drive sampler should be not less than 2 inches, and 
preferably 3 inches or more in diameter. Where loose 
sands or soft silts are encountered, a special sampler 
with a flap valve or a plunger is usually required to hold 
the material in the barrel. A bailer can be used to obtain 
sands and gravel samples from below the water table. 
Split-spoon samples should be used to obtain 
representative samples in all cases where piles are to be 
driven or the density of cohesionless materials must be 
estimated. 

4-3. Field measurements of relative density and 
consistency. 

a. Standard Penetration Test (SPT). This test 
is of practical importance as it provides a rough 
approximation of the relative density or consistency of 
foundation soils and should always be made when piles 
are to be driven. The split spoon is usually driven a total 
of 18 inches; the penetration resistance is based on the 
last 12 inches-the first 6 inches being to seat the sampler 
in undisturbed soil at the bottom of the boring. "Refusal" 
is usually taken at a blow count of 50 per 6 inches. 
(Commercial firms will usually charge an increased price 
per foot of boring when the blow count (N-value) ranges 
from greater than 50 to 60 blows per foot of penetration 
due both to reduced daily footage of drilling and wear of 
equipment.) An approximate correlation of results with 



density for cohesionless soils is shown ir l figure 4T1 and 
with (j) in Ifigure 4-2~Tl but (j> values above 35 degrees 
should not be used for design on the basis of these 
correlations. There is no unique relationship between re- 
values and relative density (DR) that is valid for all 
sands. The SPT data should be correlated with tests on 
undisturbed samples on large projects. 

b. Cone penetration tests. In this test, a cone- 
shaped penetrometer is pushed into the soil at a slow 
constant rate; the pressure required to advance the cone 
is termed the penetration resistance. The Dutch cone is 
the most popular. The penetration resistance had been 
correlated with relative density of sands and undrained 
shear strength of clays. 

c. Vane tests. The in situ shear strength of 
soft to medium clays can be measured by pushing a 
small four-blade vane, attached to the end of a rod, into 
the soil and measuring the maximum torque necessary 
to start rotation (shearing of a cylinder of soil of 
approximately the dimensions of the vane blades). The 
undrained shear strength, s u , is computed from this 
torque, T, as follows: 



s u n 



( d'h + 
V 2 



(4-4) 



O.W.L. it 10* 



o.w.l. «t ncr 



?' 


»."• 


V 


T 


*, ' 


r 


v 


•p 


•0' 


r 


?' 


10" 




f 


1°' 


r 


r 


7p' 


10' 


r 



* 10 

(J ,w 



































vemr 


oemt 


















































































0C» 


se 















































mcotu* 
















Loose 

— H 


_____ =u 


















i 

vemr loose 

i 






3 






1 


t 




i 


r 


i 



ISO 

140 
ISO 
110 
110 
100 



TO 
•O 

_ _ (. g 

40 
90 

to 



VERTICAL EFFECTIVE STRESS AT SAMPLE LOCATION - KSF 



(Courtesy of the American Society of Civil Engineers . "Task 
Committee for Foundation Design Manual of the Committee on 
Shallow Foundations " Journal, Soil Mechanics and 
Foundations Division , No. SM6. 1972.) 



Figure 4-1. Relative density of sand from the standard penetration test. 



4-5 



TM 5-818-1 /AFM 88-3, Chap. 7 



where 

d = diameter of vanes 

h = height of vane 

<p = 2/3 for uniform end-shear (usual 

assumption) distribution 

3/5 for parabolic end-shear 

distribution 

1/2 for triangular end-shear 

distribution 
The vane shear is best adapted to normally consolidated, 
sensitive clays having an undrained shear strength of 
less than 500 pounds per square foot. The device is not 
suitable for use in soils containing sand layers, many 
pebbles, or fibrous organic material. Vane tests should 
be correlated with unconfined compression tests before 
they are used extensively in any area. Strength values 
measured using field vane shear tests should be 
corrected for the effects of anisotropy and strain rate 
using Bjerrum's correction factor, X, shown in l figure 4^31 
This value represents an average and should be 
multiplied by 0.8 to obtain a lower limit. The correction is 
based upon field failures. 

d. Borehole pressuremeter test. A 

pressuremeter can be used to obtain the in situ shear 
modulus and/or K . Several versions of the device exist 
including self-boring equipment, which tends to avoid the 
loss of K conditions caused by soil relaxation when a 



hole is pre-drilled and then the device is inserted. The 
method is subject to wide interpretation and should not 
normally be employed in conventional investigations. 

4-4. Boring logs. The results of the boring program 
shall be shown in terms of graphic logs of boring. The 
logs of borings shall be prepared in accordance with 
g overnmenta l standards. A typical log of boring is shown 
in lfiqure 4-4| 

4-5. Groundwater observations. In many types of 
construction it is necessary to know the position of the 
groundwater level, its seasonal variations, how it is 
affected by tides, adjacent rivers or canals, or the water 
pressures in pervious strata at various depths. Possible 
future changes in groundwater conditions, such as those 
resulting from irrigation or reservoir construction, should 
be anticipated. 

a. Boreholes. With many fine-grained soils it 
may be necessary to wait for long time periods before 
water table equilibrium is reached in boreholes. 
Observations made in a borehole during or shortly after 
drilling may be misleading. Even with pervious soil, a 
water level 



60 



50 



w 

D 
J 
< 40 



30 



z 

o 
u 

s 
o 

-I 

ID 20 

I- 
Q. 
<n 



































i^^ 










%y^ 












<£, 














£ = 25 









2 3 4 

OVERBURDEN PRESSURE, KSF 



Schmertmann, 
" Proceedings, 



The Measurement of In Situ 
Conference on In Situ 



(Courtesy of J. H. 

Shear Strength, 

Measurement of Soil Properties . Raleigh. N. C, Vol 2, 1975, 

pp 57-180. Reprinted by permission of the American Society 

of Civil Engineers. New York.) 



Figure 4-2. Rough correlation between effective friction angle, standard blow count, and effective overburden pressure. 



4-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



reading should be taken 24 hours or more after drilling is 
stopped. Water level readings obtained in drill holes 
should be shown on the boring log with the date of the 
reading and the date when the drill hole was made. 

b. Piezometers. Piezometers provide an 
accurate means for determining the groundwater level 
over a period of time. In pervious strata, a temporary 
piezometer may consist of a section of riser pipe, the 
open bottom end of which is placed in a bag (filter) of 
coarse sand or gravel. The annular space between the 
piezometer riser pipe and the drill hole immediately 
above the stratum in which the water level is to be 



determined should be sealed off with well-tamped clay or 
cement, or chemical grout. In granular soils where a 
more permanent system is desired, a 2-foot section of 
well-point screen can be attached to the bottom of the 
pipe. A well-point screen should be selected that will 
prevent entrance of foundation materials into the screen, 
or else a suitable filter material should be used. For all 
piezometers, seal the top several feet below ground 
surface around the riser pipe to prevent infiltration of 
surface water. In granular soils, the riser pipe is normally 
about 1 1/4-inches inside diameter and generally made 
of plastic. In cohesive soils, a Casa- 








Cs ) = yes ) 

u DESIGN u VANE 



1.2 



c 1.0 



I 
X 



P 0.8 

o 

UJ 

8 



Q6 






20 40 60 80 

PLASTICITY INDEX, PI 



100 



120 



(Courtesy of L Bjerrum, "Embankment on Soft Ground, " 
Proceedings, Conference on Performance of Earth and Earth- 
Supported Structures . Purdue University. Lafayette, Ind., 
Vol2, 1975. Reprinted by permission of the American Society 
of Civil Engineers, New York.) 



Figure 4-3. Correction factor for vane strength. 



4-7 



TM 5-818-1 / AFM 88-3, Chap. 7 



BORING 
V-27 



10 



20 



30 



40 



50 



60 



70 " 



80 



60 
- 32 

22 
58 
56 
53 
57 



I 



60 
54 
53 
51 

50 
29 
35 
37 
30 
32 
32 
31 



I 



DK. GRAY 
ORG. 

9/19/78 



GRAY 
LEAN CLAY 
GRAY 
SILT 



DK.GRAY 
FAT CLAY 



GRAY 
SILT 



GRAY 
LAMINATED" 



TAN 

SANDY 

SILT 



WATER CONTENT. PERCENT 
20 40 60 



NATURAL 
WATER CONTENT 

o 



PL 



o 
o 



o 

-o— 



o 

-o— 



80 



UNDRAINED SHEAR 
STRENGTH S U ,TSF 

0.5 1.0 



OVERBURDEN AND PRE- 
CONSOLIDATION PRESSURE. TSF 



LL 

— t 




-NOTE: 
PLOTTED POINTS 
ARE SHEAR 
STRENGTHS FROM 
QTRIAXIAL TESTS 











—OVERBURDEN 




PRESSURE 






















, 1 


o 




\ 1 


1 




\ 1 






\ 1 

\o 

v — 


\ 


NOTE: PLOTTED POINTS ARE 


PRECONSOLIDATION 


PRESSURES CALCULATED 


FROM CONSOL. TESTS 



U. S. Army Corps of Engineers 



Figure 4-4. Typical log of boring. 



4-8 



TM 5-818-1 / AFM 88-3, Chap. 7 



grande-type piezometer ffig 4-5) l is recommended. The 
water level in the piezometer is determined by means of 
a plumb line or sounding device. If the piezometric level 
is above ground surface, a manometer or a Bourdon 
gage can be connected to the riser pipe to greatly 
decrease the time for equilibrium to be achieved. If 
rapidly changing pore water pressures in clay are to be 
determined, use closed system piezometers. 

c. Field pumping tests. Where accurate 
knowledge of the permeability of the foundation soils is 
necessary, field pumping tests offer the most reliable 
means. A rough estimate of the average permeability of 
the material around the bottom of a cased drill hole may 
be obtained by lowering or raising the water level in the 
casing and observing the rise and fall of the water level 
as a function of time with respect to the stabilized 
piezometric water level (TM 5-818-5/AFM 88-5. IChapterl 
[6] para 38). A field pumping test is best per- formed by 
pumping from a well in which a constant flow is 
maintained until the drawdown has stabilized, and when 
groundwater levels are measured at several remote 
borings or piezometers. It is desirable that well screens 
fully penetrate the strata for which the permeability is to 
be measured. Formulas for computing the overall 
permeability of a pervious stratum exh ibiting gravity or 
artesian flow are shown in Ifiqure 3-571 The formula for 
the special case of a fully penet rating well and artesian 
aquifer is given as an example in lfigure 4 : 6~1 Methods for 



performing field pumping te sts are described in TM 5- 
818-5/AFM 88-5, 1 Chapter 6] 

4-6. In situ load tests. Load tests are commonly 
made on test piles to confirm design capacity and may 
occasionally be used to determine bearing capacity and 
settlement characteristics. In general, specialized 
equipment and procedures are required to perform load 
tests and considerable judgment and expertise must be 
employed to interpret results. Plate load tests are 
occasionally used for bearing capacity determinations. 

4-7. Geophysical exploration. Geophysical 

methods of subsurface exploration are well suit ed for 
large sites due to the increasing cost of borings. ITablel 
1 4-4 I summarizes those geophysical methods most 
appropriate for site exploration. These methods are 
useful for interpolation between borings. Geophysical 
data must be used in conjunction with borings and 
interpreted by qualified experienced personnel, or 
misleading information is almost certain to result. The 
two most applicable geophysical methods for exploring 
foundations currently in use are seismic refraction and 
electrical resistivity. Information secured by seismic 
refraction is primarily depth to bedrock and delineation of 
interfaces between zones of differing velocities. An 
electrical resistivity survey is superior in differ- 



APPROX 1 



Pr 



3/8-IN. ID SARAN STANDPIPE 



RUBBER BUSHING 



IMPERVIOUS SEAL- 



!i. 



urn 
Ml' i 

ru)-i 



- NORTON POROUS TUBE 



IS' 



■ RUBBER STOPPER 




1-IN. TO 2-IN. DIAM 
STEEL STANDPIPE - 



COUPLING OR REDUCER 

2-IN. TO 3-IN. DIAM, 
2-FT TO 4-FT LONG 
STANDARD WELLPOINT 
SCREEN 



— e 



FILTER SAND - 



BOREHOLE, AT LEAST 
1-IN. ANNULAR SPACE 
AROUND WELLPOINT . 



5 



a. CASAGRANDE PIEZOMETER b. WELLPOINT 

U. S. Army Corps of Engineers 

Figure 4-5. Typical details of Casagrande piezometer and piezometer using uwell screen. 

4-9 



TM 5-818-1 / AFM 88-3, Chap. 7 




DISTANCE FBOM WELL IN FEET (LOG SCALE) 

NOTE. WELL SHOULD BE PUMPED AT A CONSTANT 
RATE OF FLOW UNTIL THE RATE OF DRAWDOWN 
IN WELL AND PIEZOMETERS IS ESSENTIALLY 
CONSTANT. 

DRAWDOWN AT EQUILIBRIUM VERSUS 
DISTANCE FROM WELL PLOTS AS A STRAIGHT 
LINE ON SEMI- LOGARITHMIC PLOT. 

Figure 4-6. Determination of permeability from field pumping test on a fully penetrating well in an artesian aquifer. 



4-10 



TM 5-818-1 / AFM 88-3, Chap. 7 



Name of Method 



Seismic methods: 



Refraction 



Continuous vibration 



Table 4-4. Surface Geophysical Methods 
Procedure or Principle Utilized 



Based on time required for seismic waves to 
travel from source of blast to point on ground 
surface, as measured by geophones spaced at 
intervals on a line at the surface. Refraction of 
seismic waves at the interface between 
different strata gives a pattern of arrival times 
vs distance at a line of geophones. 

The travel time of transverse or shear waves 
generated by a mechanical vibrator consisting 
of a pair of eccentrically weighted disks is 
recorded by seismic detectors placed at 
specific distances from the vibrator. 



Applicability 



Utilized to determine depth to rock or other 
lower stratum substantially different in wave 
velocity than the overlying material. Used only 
where wave velocity in successive layers 
becomes greater with depth. Used to determine 
rock type, rock and soil stratification, depth of 
weathered zone, etc. 

Velocity of wave travel end natural period of 
vibration gives some indication of soil type. 
Travel time plotted as a function of distance 
indicates depths or thicknesses of surface 
strata. Useful in determining dynamic modulus 
of subgrade reaction and obtaining information 
on the natural period of vibration for design of 
foundations of vibrating structures. 



Electrical methods: 



Resistivity 



Drop in potential 



Acoustic method 



Based on the difference in electrical 
conductivity or resistivity of strata. Resistivity of 
subsoils at various depths is determined by 
measuring the potential drop and current 
flowing between two current and two potential 
electrodes from a battery source. Rerirstivity 
irs correlated to material type. 



Based on the determination of the ratio of 
potential drops between three potential 
electrodes as a function of the current imposed 
on two current electrodes. 



The time of travel of sound waves reflected 

from the mud line beneath a body of water and 

a lower rock surface is computed by 

predetermining the velocity of sound in the 

various media. 



Used to determine horizontal extent and depth 
of subsurface strata. Principal applications for 
investigating foundations of dams and other 
large structures, particularly in exploring 
granular river channel deposits or bedrock 
surfaces, sources of construction material, 
potential infiltration and seepage zones, and in 
cavity detection. 

Similar to resistivity methods but gives sharper 
indication of vertical or steeply inclined 
boundaries and more accurate depth 
determinations. More susceptible than 

resistivity method to surface interference and 
minor irregularities in surface soils. 

Currently used in shallow underwater 
exploration to determine position of mud line 
and depth to hard stratum underlying mud. 
Method has been used in water depths greater 
than 100 feet with penetrations of 850 feet to 
bedrock. Excellent display of subsurface 
stratification. Used most efficiently in water 
depths up to 50 feet with penetrations of 
additional 350 feet to bedrock. 



Note: Instrumentation for methods listed above currently available at WES. May be furnished as a service to Districts 
and Divisions on request. Table adapted from "Design Manual, Soil Mechanics, Foundations, and Earth 
Structures, " Department of the Navy. Bureau of Yards and Docks. 



4-11 



TM 5-818-1 / AFM 88-3, Chap. 7 



entiating between sand and clays. Both methods require 
distinct differences in properties of foundation strata 
materials to be effective. The resistivity method requires 
a high-resistivity contrast between materials being 
located. The seismic method requires that the contrast 
in wave transmission velocities be high and that any 
underlying stratum transmit waves at a higher velocity 
(more dense) than the overlying stratum. Some 
difficulties arise in the use of the seismic method if the 
surface terrain and/or layer interfaces are steeply sloping 
or irregular instead of relatively horizontal and smooth. 



4-8. Borehole surveying. 

a. Downhole surveying devices can be used in 
correlating subsurface soil and rock stratification and in 
providing quantitative engineering parameters, such as 
porosity, density, water content, and moduli. Once a 
boring has been made, the cost of using these tools in 
the borehole is relatively mode st. Differ ent devices 
currently in use are summarized in | table 4-5] 

b. These devices can allow cost savings to be 
made in the exploration program without lessening the 
quality of the information obtained. 



4-12 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 4-5. Borehole Surveying Devices 



Device 3 



Electric logging 

Spontaneous potential 
(SP) 



Single-point 
resistivity 



Multiple-point 
resistivity 

Radiation logging 
Gamma 



Neutron 



Gamma-gamma 



Sonic logging 
Acoustic 
velocity 



Measurement Obtained 

Natural voltages between 
fluids in materials of 
dissimilar lithology 

Resistance of rock adja- 
cent to hole 



Resistivity of 
formations 



Natural gamma radiation 
of materials 



Hydrogen atom 
concentration 



Gamma radiation 
absorption 



Travel time of primary 
and shear wave 
velocities 



(Continued) 



Primary Use 



Differentiating be- 
tween sands and 
clays 

With SP log provides 
good indication of 
subsurface strati- 
fication and soil 
type 

Determination of mud 
infiltration and 
effective porosity 

Identification of 

clay seams, loca- 
tion of radioactive 
tracers, and with 
SP and resistivity 
logs provides in- 
formation on rela- 
tive porosity 

Determination of 
moisture content 
and porosity below 
zone of saturation 
porosity 

Correlates with bulk 
density and useful 
to determine poros- 
ity if grain spe- 
cific gravity is 
known 

With caliper and den- 
sity logs determine 
dynamic elastic and 
shear moduli of in 
situ rock 



1 All devices currently available at WES. May be furnished as a service to Districts and Divisions on request. (Sheet 1 of 3) 



4-13 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 4-5. Borehole Surveying Devices-Continued 



Device 3 



Sonic logging 
(Continued) 

Acoustic 
Imagery 



Fluid logging 
Temperature 



Fluid resistivity 



Trace ejector 
Fluid sampler 

Visual logging 
Borescope 



Borehole camera 



Measurement Obtained 



Reflected acoustic 
energy 



Temperature gradient 
in borehole 



Electrical resistance of 
borehole fluids 



Controlled ejection of 
trace elements 

Borehole or formation 
fluid from predeter- 
minded depths 



Visual image of the 
sidewalls of a bore- 
hole to depths of 100 
ft or less-a peri- 
scopic instrument 

Photograph of a 360 deg 
sweep of the borehole 
wall taken approxi- 
mately at right angles 
to the wall. Expo- 
sures timed so that 
slight overlap of each 
photograph is obtained 



Primary Use 



Locate fractures and 
voids and strike 
and dip of joints, 
faults, bedding 
planes, etc. 



Determine geothermal 
gradient and defi- 
nition of aquifers 

With temperature data 
the determination 
of dissolved 
solids-locate 
zones of water loss 
or gain 

Groundwater flow 
patterns 

Uncontaminated sam- 
ples for water 
quality studies 



Cheap, rapid examina- 
tion of borehole 
walls 



Examination of bore- 
hole conditions, 
bedding, joints, 
etc. 



(Continued) 



(Sheet 2 of 3) 



4-14 



TM 5-818-1 / AFM 88-3. Chap. 7 



Table 4-5. Borehoe Surveying Devices-Continued 



Device 3 



Measurement Obtained 



Primary Use 



Visual logging 
(Continued) 

Downhole camera 



Borehole television 



Photograph of the side- 
walls of the borehole 
taken from the bottom 
of the camera. Several 
feet of hole below de- 
vice taken with each 
exposure 

Image of the sidewalls 
of the borehole dis- 
played at the time of 
exposure on a surface 
monitor 



Examination of bore- 
hole conditions, 

bedding, joints, 
etc. 



Rapid examination of 
borehole conditions 
as the device is 
lowered or raised 



Miscellaneous logging 
Caliper 



Borehole surveyor 



Borehole size 



Downhole directional 
survey 



Washouts, fractures, 
etc., needed for 
interpretation of 
other logs 

Precise location and 
attitude of fea- 
tures recorded on 
other logging tools 



U. S. Army Corps of Engineers 



(Sheet 3 of 3) 



4-15 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 5 
SETTLEMENT ANALYSES 



5-1. Settlement problems. 

a. Significant aspects of the settlement of 
structures are total settlement-magnitude of downward 
movements- and differential settlement-differe nce in 
sett lements at different locations in the structure. ITablel 

15-1 M ists conditions that cause settlements which occur 
during construction and result in only minor problems 
and postconstruction settlements which occur after a 
structure is completed or after critical features are 
completed. Differential settlements distort a structure. A 
structure can generally tolerate large uniform, or nearly 
uniform, settlements. 

b. Differential settlement can have a number 
of undesirable results: 

(1) Tilting is unsightly. A tilt of 1/250 can 
be distinguished by the unaided eye. 

(2) Moderate differential settlement 
causes cracking and architectural damage. With 
increasing differential settlement, doors and windows 
may become distorted and not open and close properly. 
Larger differential settlements may cause floors and 
stairways to become uneven and treacherous and 
windows to shatter. At this point, the usefulness of the 
building has been seriously impaired. 

(3) Severe differential settlements may 
impair structural integrity and make structures 
susceptible to collapse during an earthquake or other 
major vibration. 

(4) If a structure settles relative to the 
surrounding ground, or the ground settles relative to the 
structure, entryways may be disrupted, and utility lines 
may be damaged where they enter the structure. 

c. Even if settlements are uniform or nearly 
so, large total settlements can also result in problems: 

(1) Sites located near a river, lake, or 
ocean may flood during periods of high water. 

(2) Surface drainage may be disrupted. If 
water ponds around and beneath structures, they may 
become inaccessible and subject to mildew and wood 
rot. 

d. Experience provides a basis for estimating 
the magnitudes of differential settlement that cause 
cracking of architectural finishes, such as plaster, 
stucco, and brick facing. Differential settlement is best 
expressed as the angular distortion in radians between 
two points. Angular distortion, which always 
accompanies settlement of a building, is determined by 
the uniformity of foundation soils, the stiffness of the 



structure and its foundation, and the distribution of load 
within the building. In conventional settlement analyses 
of the type described in this manual, the stiffness of the 
building and foundation are no t consider ed. Tolerable 
angular distortions are listed in liable 5-"2l and empirical 
correlations that may be used to estimate probable 
angular distortions based on calculated maximum 
settlements are summarized in table 5-3.1 Because of 
the natural variability of soils, differential settlement will 
occur though total settlements are calculated to be 
uniform. An indirect means for controlling differential 
settlement is to limit total settlement to 3 inches for 
structures on clay and to 1 1/2 inches for structures on 
sand. 

5-2. Loads causing settlement. 

a. Loads causing settlement always include 
the estimated dead load and a portion or all of the live 
load. For office buildings, about 50 percent of the 
estimated building live load may be assumed to cause 
settlement. For heavily loaded warehouses and similar 
structures, the full live load should be used. 

b. For many purposes, settlements need be 
computed only for the maximum dead load plus 
settlement-causing live load. Occasionally, settlements 
occurring during a part of the construction period must 
be computed. This may require additional stress 
computations for partial loading conditions. 

c. Loads that are less than the 
preconsolidation stress cause minor settlements 
because only recompression of soil occurs. The 
increment of loading that exceeds the preconsolidation 
stress causes relatively large settlements and occurs 
along the virgin compression portion of laboratory 
consolidation curves. A careful estimate of 
preconsolidation stresses is essential for settlement 
a nalyses. M eans for estimating such stresses are given 
in |chapter~3l 

5-3. Stress computations. 

a. One of the first steps in a settlement 
analysis is computation of effective overburden stresses 
in the soil before and after loading. The initial stress p/ 
at any depth is equal to the effective weight of overlying 
soils and may be determined by multiplying the effective 
unit weight of the soil by its thickness. It is customary to 
construct a load-depth diagram by plotting 



5-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 5- 1 . Causes of Settlement 



Cause 



Comment 



Compression of foundation 
soils under static loads. 



Compression of soft clays 
due to lowering ground- 
water table. 



Soft, normally consolidated clays and 
peaty soils are most compressible. 
Loose silts, sands, and gravels are 
also quite compressible. 

Increased effective stress causes 
settlement with no increase in sur- 
face load. 



Compression of cohesionless 
soils due to vibrations. 



Loose sands and gravels are most sus- 
ceptible. Settlement can be caused by 
machine vibrations, earthquakes, and 
blasts. 



Compression of foundation 
soil due to wetting. 



Shrinkage of cohesive soils 
caused by drying. 



Loss of foundation support 
due to erosion. 

Loss of foundation support 
due to excavation of 
adjacent ground. 

Loss of support due to 
lateral shifting of the 
adjacent ground 

Loss of support due to 
formation of sinkhole. 

Loss of support due to thaw- 
ing of permafrost.foundation heat. 

Loss of support due to 
partial or complete 
liquefaction. 

Downdrag on piles driven 
through soft clay. 



Loose silty sands and gravels are most 
susceptible. Settlements can be 
caused by rise in groundwater table 
or by infiltration. 

Highly plastic clays are most suscept- 
ible. Increase in temperature under 
buildings containing ovens or fur- 
naces may accelerate drying. Wetting 
of highly plastic clays can cause 
swelling and heave of foundations. 

Waterfront foundations must extend below 
maximum erosion depth. 

Most pronounced in soft, saturated clays. 



Lateral shifting may result from land- 
slides, slow downhill creep, or move- 
ment of retaining structures. 

Soils overlying cavernous limestone and 
broken conduits are susceptible. 

Permafrost should be insulated from 



Loose, saturated sands are most 
susceptible. 



Loading on piles is increased by nega- 
tive skin friction if soil around 
upper part of pile settles. 



U. S. Army Corps of Engineers 



5-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



stress versus depth, using average unit moist weights of 
soil above the water table and average unit submerged 
weights below the water table. 

b. The final stress p at any depth is equal to 
the effective overburden stress after the structure is 
completed plus the stress resulting from the structure 
load. If the structure is founded on individual footings, 
the final stress is the sum of stresses imposed by all 
footings. 



c. Foundation stresses caused by applied 
loads are generally computed assuming the foundation 
to consist of an elastic, isotropic, homogeneous mass of 
semiinfinite extent, i.e., the Boussinesq case. The 
increment of stress at various depths is de termined by 
means of influence values, such as shown in l figures 5-11 
I and 5-2,1 which give the vertical stress beneath a 
rectangular area for uniform and triangular distributions 
of load, respectively. Influence values for vertical 



Table 5-2. Value of Angular Distortion (8/&) That Can Be Tolerated Without Cracking 



IRREGULAR SETTLEMENT 



REGULAR SETTLEMENT 





Type of Building 



L/H 



Allowable 
W 



Steel frame with flexible siding 



0.008 



Steel or reinforced concrete frame 
with insensitive finish such as 
dry wall, glass, or moveable 
panels 



0.002 to 
0.003 



Steel or reinforced concrete frame 
with brick, block, plaster, or 
stucco finish 



>5 
<3 



0.002 
0.001 



Load-bearing brick, tile, or 
concrete block walls 



>5 
<3 



0.0008 
0.0004 



Circular steel tanks on flexible 
base, with fixed top 



0.008 



Circular steel tanks on flexible 
base, with floating top 



0.002 to 
0.003 



Tall slender structures, such as 
stacks, silos, and water tanks, 
with rigid mat foundations 



0.002 



U. S. Army Corps of Engineers 



5-3 



TM 5-818-1 / AFM 88-3, Chap. 7 



stress beneath a circular area are shown i ri figure S 1 ^ If 
the foundation consists of a large number of individual 
footings, influence charts based on the Boussinesq case 
will greatly facilitate the computation of stresses. 
Programs for digital computers and programmable 
calculators are also available. 

d. A structure excavation reduces stress in 
foundation subsoils. The decrease in vertical stresses 
caused by the weight of excavated material is computed 
in the manner described in the previous paragraph. The 
bottom of the excavation is used as a reference; vertical 
stresses produced by the weight of excavated material 
are subtracted algebraically from the original overburden 
pressure to compute final foundation stresses. 

5-4. Settlement of foundations on clay. 

a. When a load is applied over a limited area 
on clay, some settlement occurs immediately. This 
immediate settlement, AH|, , has two components: that 
caused by distortion or change of shape of the clay 
beneath the loaded area, and that caused by immediate 
volume change in unsaturated soils. In saturated clays, 
there is little or no immediate volume change because 
time is required for water to drain from the clay. 

b. Immediate settlemen ts can be estimated 
using methods given in lchapterTQ ). Values of undrained 
modulus determined from the slopes of stress-strain 



curves from unconsolidated-undrained laboratory 
compression tests are frequently only one-half or one- 
third as large as the in situ modulus. This difference is 
due to disturbance effects, and the disparity may be 
even more significant if the amount of disturbance is 
unusually large. T he undrained modulus of the clay may 
be estimated from l figure 3-20.1 The values of the K in 
this figure were determined from the field measurements 
and, therefore, are considered to be unaffected by 
disturbance. The value of Poisson's ratio is equal to 0.5 
for saturated clays. For partly saturated clays, a value of 
0.3 can be assumed. Because immediate settlements 
occur as load is applied and are at least partially included 
in results of laboratory consolidation tests, they are often 
not computed and only consolidation settlements are 
considered to affect a structure. 

5-5. Consolidation settlement. Consolidation 
settlement of cohesive soil is normally computed from 
pressure-void ratio relations from laboratory 
consolidation tests on representative samples. Typical 
examples of pressure-void ratio curves for insensitive 
and sensitive, normally loaded c lays, and 
preconsolidated clays are shown in | figure 3^71 
Excavation results in a rebound of foundation soils and 
subsequent recompression when structure loads are 
added. This 



Table 5-3. Empirical Correlations Between Maximum (A) and Angular Distortion (5/&J 



Type of Foundation 
Mats on sand 
Rectangular mats on varved silt 

Square mats on varved silt 
(0.0005 to 0.0003) 

Mats on clay 

Spread footings on sand 

Spread footings on varved silt 

Spread footings on clay 



Approximate Value 

of 5A& for a 

A= 1 in. 

1/750(0.0013) 

1/1000 to 1/2000 
(0.001 to 0.0005) 

1/2000 to 1/3000 



1/1250(0.0008) 
1/600(0.0017) 
1/600(0.0017) 
1/1000(0.0010) 



a 5A3- increases roughly in proportion with A. For A = 2 in., values of 5A3- would be about 
twice as large as shown, for A = 3 inches, three times as large, etc. 

(Courtesy of J. P. Gould and J. D. Parsons, "Long - Term Performance of Tall Buildings of New York City 
Varied Silts, " Proceedings, International Conference on Planning and Design of Tall Buildings , Lehigh 
University, Bethlehem, Pa., 1975. Reprinted by permission of American Society of Civil Engineers, New 
York.) 

5-4 



sequence should be simulated in consolidation tests by 
loading the specimen to the existing overburden 
pressure po, unloading to the estimated stress after 
excavation p ex c, and reloading the specimen to define the 
p-e curve at loads in excess of overburden and 



TM 5-818-1 / AFM 88-3, Chap. 7 

preconsolid ation stresses. Curves designated K u in 
I figure 3-7~l are laboratory p-e curves. Soil disturbance 
during sampling affects laboratory p-e curves so that it 
usually 



0.26 



0.26 



0.24 



0.22 



0.20 



O.I 6 



0.16 



t 0.14 



0.12 



0.10 



008 



0.06 



0.04 



0.02 



0.00 



1 0.28 



0.24 



0.20 



0.16 




U. S. Army Corps of Engineers 

Figure 5- 1 . Vertical stress beneath a uniformly loaded rectangular area. 



5-5 



TM 5-818-1 AFM 88-3, Chap. 7 




■"JL. 












I 








! : i 1 11 


: : 3^.; 
















VALUES O 


n j Tttt' 


: ^ : .. 














^^ 


r^ 


-• 




r «I + ..TI 


rH- 












L ' ^** T- 4* *"* "" 








V ! I 


^i::: 




;(:: 
















Sz:::i:J 


: = :/. 










.^•'-;-r . 










AiftS' 


V 










fer;. 












^ 1 








^ 




* ** 


■-"t **"" 












~ i 1 


r^j: 




&■**" 


^~ 














if ,ttt"T' 


■ "' "T " 














«^2: 


•"*._ 


aIU 


""" 






] .iliiiiii 












L-ilil 



:~ffi 


■•.33il 


!!: 














-3 




Iff! 












1 


-;rj 


















:;;lli 




VK.UMOP 


^ * '■' 






v.5~-: 




II. 

I 1 ' 


















TT~1 






I 
it 




' ■ I ■ 


: '" 




1 
ii 




V 












'- j 


:1 4 






!• 




+4 


f::. 


„ui;:- : | 


t;; 










.''■'■t 






4'- 


- — 


— 


1* 

-r 





....1. 


:■'- 


"-'.. 


If!' 






- 




* ■ •* ,\ 


j>*"" i 




















• 
• 




T[ 


;::§: 










■"* 


** 


1' •■' ■ + 


<~ ^ 


_— 








— 
i 


..... 


" \ 


-S5r : 


•W 








■-""T 


7~" 














i 




1T"I 




"T-* 




= ..- + 


nfr-^i 


__ 


— 


= 


* j 

■T" 

-»- 

1 




=:;| 



vemcM. MtML natu awcn w mmm. ■nun comb b 



VAUIUOVb 



vorriou. mouml irant ■ 



P/gyre 5-2. Vertical stress beneath a triangular distribution of load on a rectangular area. 



5-6 



TM 5-818-1 AFM 88-3, Chap. 7 



becomes necessary to construct a field p-e curve 
simulating consolid ation in the f ield. These constructions 
are also shown in Ifiqure 3-7~l They are based on the 
assumption that the straight lower branch of the field p-e 
curve, K, intersects the laboratory curve at e = 0.4e . 
Furthermore, the field curve must pass through point a, 
corresponding to the present overburden pressure and 
the natural void ratio. The field compression index, C c , is 
taken as the slope of the straight lower branch of K on 
the semilogarithmic diagram. 

a. Settlement computations. The total 
settlement, AH, of a foundation stratum is computed 
according to the following formula: 



where 



AH = 



ei, e 2 



eo 
H 



e_i_; 

1 + 



_e_2 

e 



(5-1) 



initial and final void ratios, 
respectively, from the field 
pressure-void curves, 

corresponding to the initial and 
final effective foundation 

pressures 

average initial void of the stratum 
total thickness of the compressible 
stratum This formula also may be 
used to estimate foundation 
rebound due to excavation. When 
the lower portion of 



VALUE OF 




O.OOI 



0.01* 
INFLUENCE VALUE, I 



Figure 5-3. Influence value for vertical stress under a uniformly loaded circular area. 



5-7 



TM 5-818-1 / AFM 88-3, Chap. 7 



the pressure void ratio curve is fairly straight, it may be 
convenient to work with the compression index, in which 
case the formula for settlement is as follows: 



AH = 



1 +e 



H logio p 2 ' (5-2) 



An example of a settlement analysis in which the 
rebound of the foundation and subsequent 
recompres sion under the building load are determined is 
shown in I figure 5-4 1 for a normally consolidated 
foundation. 

b. Rate of settlement. The rate of settlement 
is determined by means of the theory of consolidation. 
This 



ground surface 








CL 

1 


R = L ; 34 

= ?RM"IENT LL 

till 












■-'OOLPCL'FT 


1 

t 


]\ 


MAC*FILL f 


BACKF 


LL 




FILL 
GROUND 1AT;B TA9LF _ 
' ' ; 37SL D LU FT 


2 T SO FT 


, 


_s 


SOFT r LftY 






\l 


/ 







'■ = 50 i-^ Cu 

. # * a 9*7 



STRATUV I 



SOFT CLflV 

V - 50 LB CU FT 

e = 0.991 



1_. 



BOCK 
GENERALIZED SOIL COKPITIOKS 



PRESSURE IN TONS PER SQ FT 

I 2 



^ NOTE \P»» ( INO \P COMPUTEO BY 

\ PRESSURE 

V 'fTEft LCWEW'NG OF 


V"..c\ 






,\\ 


/ 
, 1 





FOUNDATION PRESSURES 



lol COMPUTATION OF OVER^UROFw PRESSURES (PJ AMD (P q ) AT MIDPOINTS OF STRATA 





INITIAL 0VHO3 




STRATUM I 


6 


S-.-.T «.T 


* 


355 ' c a " ' " IT 


8 


Ji .».«" 



DEN PRESSURE IP ' 

STRATUM n 



RBURDEN PRESSURE AF 



TER TABLE iS LOWERED 



'00 

" iooo 


300 T 


53 FT 


2- iH 
20CO 


03fl T 


SO FT 




475 T 


50 FT 



»-"— . >■>•&&) 



STRATUM D 






* \ 2000 / " 



P„ 518 T S3 FT » 813 T SQ FT 

(M COMPUTATION OF PRESSURE AFTER FXCAVAT10N (P„ c ) AT MID-POINT OF STRATA 



STRATUM n 



'RES5URE BEFORE EXCAVATION fP^I - 0.725 T W FT 

<RES5uRE RELEASE DUE TO EXCAVATION I \P J = -0 620 T 53 FT 



(c) COMPUTATION OF FINAL PRESSURE (P,J AT MID-POI^T OF STRATA 





NET LOAD AT BASE OF STRUCTURE 












Ol - PERMANENT LL - 1 000 T SO FT 












SOIL REMOVED , 












8 ■ 
NET STRl 


too 1 

Si — 

2000 ] 












CTURE 






LOAD 


- ■ 375 T SO FT 










tdj 


COMPUTATION OF 


FOUNDATION REBOUND 














STRATUM 

THICKNESS P a P «M 








REBOUND 

"f " ") 




STRATUM 


H T S3 FT T SQ FT 


9*7 


96* 


-0 017 


V."«J 




I 


» FT 5* OH 


OCT FT 




n 


14 FT ei 49 


991 


999 


-0 COS 


06 FT 



TOTAL REBOUND : D 13 F 



It) COMPUTATION OF FOUNDATION SETTLEMENT 



T SO FT T SO FT ••« «i \e V 1 * "a/ 



INITIAL OVERBURDEN PRESSURE tF» J = 0,538 T SQ f 
INCREASE IN FOUNDATION PRESSURE 

DDE TO NET STRUCTURE LOA3 l\P) ■ I.3S6 T SQ F 

P ] = I 894 T SQ F 

NOTE: \" " NET STRUCTURE LOAD ' * 
*„ OBTAINED FROM FIG 5-1 




TOTAL SETTLEMENT - 1.61 FT 



PRESSURE 'LOG SCALE) 

PRESSURE - VOID RATIO CURVES 



U. S. Army Corps of engineers 



Figure 5-4. Example of settlement analysis. 
5-8 



theory relates the degree of consolidation and time 
subsequent to loading according to the following 
expression: 



with 



where 
U = 



c v = 



U(%) = f(T) 



Cv 



(5-3) 
(5-4) 



IT 



degree of consolidation or ratio of 
settlement that has occurred at a given 
time to the ultimate settlement 
dimensionless number called the time 
factor that depends upon loading and 
boundary conditions 
property of the soil known as the 
coefficient of consolidation 



t 



TM 5-818-1 / AFM 88-3, Chap. 7 

length of the drainage path, which in 

the case of a specimen or stratum 

draining from top and bottom would be 

half the thickness of the specimen or 

stratum 

time corresponding to U 



The relation between time factor and percent 
consolidatio n for various boundary conditions is shown in 
I figure 5-5l If the values of c v and H are known for a 
stratum of clay with given boundary conditions, the 
theoretical curve can be replotted in the form of a 
percent consolidation-time curve; if the ultimate 
settlement of the layer has been computed, the curve 
can be further modified into a settlement-time curve. In 
order to compute, c v , it is necessary to transform the 
laboratory time-consolidation curve for the load 
increment in question into the theoretical curve. A 
method for ad- justing the laboratory curve in order to 
compute the 




0.10 1.00 10 

TIME FACTOR -T 

ONE DIMENSIONAL CONSOLIDATION -CONSTANT LOADING 



DOTATIONS 



% , _S i ft = BATE OF LOADING 



•Krone 

TIMC 1. 


"^/C 


TIME FACTO* AT t 
ITWO-WAr OAAtMAOEl 

\"S. 




/ AFTER 
TIME l ( 




"* 






010 
TIME FACTOR - T 



U. S. Army Corps of Engineers 



Figure 5-5. Time factors for various boundary conditions. 



5-9 



TM 5-818-1 / AFM 88-3, Chap. 7 



coefficient of consolidation, c v , is shown in l figure 3-7\ 
Thus, the actual time required for the various 
percentages of consolidation to occur in the field can be 
determined by the following formula: 



Ttf 
c v 



(5-5) 



where 
t f 

H 
T 



time for U(%) consolidation in the 

field stratum 

length of the drainage path in the 

field 

time factor corresponding to U(%) 

consolidation 



When settlement occurring during the construction 
period may be of in terest, the values of T and U also 
can be obtained frorr i figure 5-5I 

c. Secondary compression. For refined 
estimates and special purposes, settlement resulting 
from secondary compression may have to be evaluated. 
The amount AhU can be calculated as follows: 



AH S = 
where 

tsc 



C„Hlog i 



(5-6) 



to 



tp + time interval during which 
secondary compression 

settlement is to be calculated 
time to complete primary 
consolidation 



H = total thickness of compressible soil 
Other terms have been previously defined. Secondary 
compression settlements may be important where 
primary consolidation occurs rapidly, soils are highly 
plastic or organic, and allowable settlements are 
unusually small. 

5-6. Settlement of cohesionless soils. The 

permeability of cohesionless soils is usually sufficiently 
great that consolidation takes place during the 
construction period. For important projects, estimate 
settlement using consolidation tests on undisturbed 
samples or samples remolded at natural density. 
Alternately, settlements may be es timated from plate 
bearing tests described in l chapter"?] Design of footings 
on cohesionless soils, based on settlement 

considerations using the Standard Penetration Test, is 
described ir lchapterTD . 

5-7. Eliminating, reducing, or coping with 
settlement. Design techniques for ameliorating 
settlement problems are summarized in I table 5-4] 
Differential settlements beneath existing structures can 
be corrected by releveling by jacks, grouting (i.e., mud 
jacking) beneath slab foundations, or underpinning. 
These techniques are expensive, to varying degrees, 
and require specialists. 



5-10 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 5-4. Methods of Eliminating, Reducing, or Coping With Settlements. 



Method 



Comment 



Use of piles, piers, or 
deep footings. 



Excavate soft soil and 
replace with clean 
granular fill. 

Displace soft soil with 
weight of granular 
fill or by blasting. 

Reduce net load by 
excavation. 



Surcharge or preload 
site before 
construction. 



Delay construction of 
buildings to be built 
on fills. 



Use a stiff foundation 

with deep grade beams. 

Install leveling jacks 

between the foundation 
and the structure 

Select a building type 
which has a large tol- 
erance for differential 
settlement. 



Differential settlements between 
buildings and surrounding ground 
can cause problems. 

Can be very costly if the com- 
pressible layer extends to 
large depth. 

Difficult to control. Pockets of 
entrapped soft soil can cause 
large differential settlements. 

Weight of one story building is 
equal to weight of one or two 
feet of soil. 

Settlement is reduced by amount 
which occurs before construc- 
tion. Preload may be limited 
by stability considerations. 

Settlement which occurs before 
construction does not affect 
building. Fill settlement can 
be accelerated using sand 
drains. 

Can greatly reduce differential 
settlements. 

Building can be releveled 
periodically as foundation 
settles. 

Steel frames, metal siding, and 
asphalt floors can withstand 
large settlements and remain 
serviceable. 



U. S. Army Corps of Engineers 



5-11 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 6 



BEARING-CAPACITY ANALYSIS 



6-1. Bearing capacity of soils. Stresses 

transmitted by a foundation to underlying soils must not 
cause bearing-capacity failure or excessive foundation 
settlement. The design bearing pressure equals the 
ultimate bearing capacity divided by a suitable factor of 
safety. The ultimate bearing capacity is the loading 
intensity that causes failure and lateral displacement of 
foundation materials and rapid settlement. The ultimate 
bearing capacity depends on the size and shape of the 
loaded area, the depth of the loaded area below the 
ground surface, groundwater conditions, the type and 
strength of foundation materials, and the manner in 
which the load is applied . Allowable bearing pressures 
may be estimated from Itable 6-1~l on the basis of a 
description of foundation materials. Bearing-capacity 
analyses are summarized below. 

6-2. Shear strength parameters. 

a. Appropriate analyses. Bearing-capacity 
calculations assume that strength parameters for 
foundation soils are accurately known within the depth of 
influence of the footing. The depth is generally about 2 
to 4 times the footing width but is deeper if subsoils are 
highly compressible. 

(1) Cohesionless soils. Estimate §' from 
the Standard Penetration Test [table 4-51 or the cone 
penetration resistance. For conservative values, use §' = 
30 degrees. 

(2) Cohesive soils. For a short-term 
anal ysis, est imate s u , from the Standard Penetration 
Test l (table 4-5 ) or the vane shear resistance. For long- 
term loadings, estimate §' from correlations with index 
properties for normally consolidated soils. 

b. Detailed analyses. 

(1) Cohesionless soils. Determine cf>' from 
drained (S) triaxial tests on undisturbed samples from 
test pits or borings. 

(2) Cohesive soils. For a short-term 
analysis, determine s, from Q triaxial tests on 
undisturbed samples with 03 equal to overburden 
pressure. For a long-term analysis, obtain §' from 
drained direct shear (S) tests on undisturbed samples. 
For transient loadings after consolidation, obtain § and c 
parameters from consolidated-undrained (R) triaxial tests 
with pore pressure measurements on undisturbed 
samples. If the soil is dilative, the strength should be 
determined from drained S tests. 



6-3. Methods of analysis. 

a. Shallow foundations. 

(1) Groundwater level (GWL). The 
ultimate bearing capacity of shallow foundations 
subjected to vertical, eccentric load s can be com puted 
by means of theformulas shown in Ifigure 6-i~| For a 
groundwater level well below the bottom of the footing, 
use a mo ist unit soil weight in the equations given in 

Ifigure 6-fl If the groundwater level is at ground surface, 
use a submerged unit soil weight in the equations. 

(2) Intermediate groundwater levels. 
Where the groundwater level is neither at the surface nor 
so deep as not to influence the ult imate beari ng capacity, 
use graphs and equations given ir l figure 6-"2l 

(3) Eccentric or inclined footing loads. In 
practice, many structure foundations are subjected to 
horizontal thrust and bending moment in addition to 
vertical loading. The effect of these loadings is 
accounted for by substituting equivalent eccentric andlor 
inclined loads. Beari ng capacity formulas for this 
condition are shown in Ifigure 6-371 An example of the 
method for computing the ultimate bearing capa city for 
an eccentric inclined load on a footing is shown ii j figurel 
6-4. 

(4) Loading combinations and safety 
factors. The ultimate bearing capacity should be 
determined for all combinations of simultaneous 
loadings. A distinction is made between normal and 
maximum live load in bearing capacity computations. 
The normal live load is that part of the total live load that 
acts on the foundation at least once a year; the 
maximum live load acts only during the simultaneous 
occurrence of several exceptional events during the 
design life of the structure. A minimum factor of safety of 
2.0 to 3.0 is required for dead load plus normal live load, 
and 1.5 for dead load plus maximum live load. Safety 
factos selected should be based on the extent of 
subsurface investigations, reliability of estimated 
loadings, and consequences of failure. Also, high safety 
factors should be selected if settlement estimates are not 
made. In general, separate settlement analysis should 
be made. 

b. Deep foundations. Methods for computing 
the ultimate bear ing capacity o f deep foundations are 
summarized in Ifigure 6-5. I These analyses are 
applicable to the design of deep piers and pile 
foundations, as subsequently described. When the base 
of the foundation is located below the ground surface at 
a depth greater 



6-1 



TM 5-818-1 / AFM 88-3. Chap. 7 



Table 6-1. Estimates of Allowable Bearing Pressure 



(These presumed values of the allowable bearing pressure are estimates and may need alteration 
upwards or downwards. No addition has been made for the depth of embedment of the foundation. 
Reference should be made to other parts of the Manual when using this table. J 



Croup 


Types and conditions of rocks 
and soils 


Strength 
of Rock Material 


Presumed Allowable 
Rearing Pressure 

Ton /so, ft 


Remarks 




Massive igneous sod metamorphic 
rocks {granite, dlorltc, basalt, 
gneiss) la sound condition a 

Foliated metamorphic rocks 
(slate, schist) In sound 
condition*' b 


■lgh to very high 
Medium to high 


100 

30 


These values are based 
on the assumption that 
the foundations arc 
carried down to unwesther- 
ed rock. 


Rock* 


Sedimentary rocks: cemented 
shale, alltatooe, sandstone, 
lines tone without cavities, 
thoroughly cemented conglome- 
rates, all In sound condition*-' 


Medium to high 


10-40 






Compaction shale and other 
argillaceous rocks la sound 
condition*^ 


Low to Medium 


5 






Broken rocks of any kind with 
moderately close epaclng of 
discontinuities (1 ft or 
greater), except argillaceous 
rocks (shale) 




10 






Thinly bedded limestone, 
sandstones, shale 




See note c 






Heavily shattered or weather- 
ed rocks 




See note c 




Non- 
cohesive 
• oils 


Dense gravel or dense sand and 
gravel 

Compact gravel or compact sand 
and gravel 

Loos* gravel or lAose land and 
gravel 

Dense sand 

Conpact a and 

Loose aand 




2-6 

<2 
>3 
1-3 
<1 


Width of foundation (B) 
not leas than 3 ft. 
Groundwater level 
assumed to be at a depth 
not lesa than B below 
the base of the founds- 
tion. 




Very stiff to hard clays or 

heterogeneous ad x tares such as 
till 




3-6 


Cohesive soils are 
susceptible to long-term 
consolidation settlement 




Stiff clays 




1.5-3 




Cohesive 
•oils 


Firm clays 

Soft clays and silts 

Very soft clays and silts 




0.75-1.5 
<0.?5 
not applicable 




Organic 
soils 


Feat and organic soils 




not applicable 




Fill 


Fill 




not applicable 





fni—.i ro2£ ■■_;. lit icr.^ a '.low T.inor cracks it spavin.- r.ct, lers thin 3 J-'t. 

T'ie ■'!:::.•:• v-iiues :'ar seiiirier.tary or f o 1 i -i *„ » i roz'f.r, apply v.-iere the strata or foliation are level or r.c-irly 

so , aril, tl-.-LT. j.-.ly i:' the irea ?,as -u'.:ii* lateral, support. Tilted strata and their relation to nearly 

r.opes or i:*r a 7a* i-o.-ir. uhall be .i^s:::..;e.i v,y a -erron knowledgeable in this fieli of '-rork. 

To tc i;j»s3L-. by .-xanir.it iotj i:. sttu, includi:^ loudi:,.-: ^sts if necessary, by a person knowledgeable in this 

These r^-f.z -ire -ivt v; .■-.--■LI on -elcniie -jf rtr-sa, nn.1 sr. exposare to water they are apt tc soften aril 



U. S. Army Corps of Engineers 



6-2 



TM 5-818-1 / AFM 88-3, Chap. 7 




30 40 50 60 70 

BEARING CAPACITY FACTORS N, N_, AND N 

c q' y 



100 



Figure 6-1. Ultimate hearing capacity of shallow foundations under vertical, eccentric loads. 



6-3 



TM 5-818-1 / AFM 88-3, Chap. 7 



1.0 

.8 

5 








\ 


\ 














X 




0.3 \ 






♦*30°. 


//\25°a 45° 




4 






X0.2 




\| 












.2 




/ 


0.1 


^V 1 

^ l 








{**" — TO DETERMINE d 




./^EPTH OF WATER TABLE . <J ' 


> 















W10TH OF FOOTING 


' • ! 


< 


) l< 


i 2 


o ao 40 





2 4 6 8 1.0 



ANCLE OF INTERNAL FRICTION ♦. OEGREES 



DEPTH OF WATER TABLE . _d_ 
DEPTH OF FAILURE ZONE* d„ 



L* LENGTH FOOTING 

ROUGH BASE 

SURFACE FOOTING 
i »» » »k^ j/;ajj* vmrrm 



B— 



SHALLOW FOOTING 
'1 jumniin rw*i 




ASSUMED CONDITIONS'. 

1. GROUND WATER LEVEL IS 
HORIZONTAL 

2. PRESENCE OF GROUND WATER 
HAS NO EFFECT ON COHESIVE 
SOIL WITH * iO. 



CONTINUOUS FOOTING: 

SURFACE FOOTING: D = 

9ul, = cN c + ?SUB + "<7T-r S UB> f N 7 

SHALLOW FOOTING: D<B 

IFd<D 

qul," cN e + 7suB D + <TT-r S UB> d "q 
+ 0.57 SUB BN T 

IF D<d<(0 + d o ) 

^l, = <= N c + TT D N q 

B 



2 T 



+ Tsub + F^t-Tsub' 

CIRCULAR FOOTING : RADIUS = R = B/2 
SURFACE FOOTING : D = 

9 U it =1 - 3cN c + T S ub + f( Tt-Tsub ) 06rn 7 

SHALLOW FOOTING : D<2R.IFd<D 

q U i, = 1 - 3cN c + TT-TsuB D + < T T -7su B >d N q + 0.67 SUB RN 7 

IF D<d<(D + d c ) 

t 'ult =1 - 3cN c + TT DN q + TsuB + F'Tt-TsuB* 0.6 RN^ 



RECTANGULAR FOOTING : 
SURFACE FOOTING : D-0 

"u.t =cN c (1+0 - 3 °> + ■ysuB + F <T-7suB> 04B "y 
SHALLOW FOOTING : D<B. IFd<D 

Pult =cN c< 1+ 0- 3 ^' + TsUB + <TT-TsuB» d N q 

+ ° i »'y SUB BN 7 
IF D<d<(D + d ) 
q ult =cN c (1+0.3°)+7 T DN q 

+ r S UB + F <TT-TsUB> 04B N 7 



U. S. Army Corps of Engineers 



Figure 6-2. Ultimate bearing capacity with groundwater effect. 



6-4 



TM 5-818-1 / AFM 88-3, Chap. 7 



FORWARO ECCENTRICITY: 



BACKWARO ECCENTRICITY: 




^ 



EFFECTIVE DIMENSIONS OF FOOTINGS UMOCn CCCCHTmC LOAO* 

CIRCULAR FOOTING: 



STRtP FOOTIMO: 



ZHJ 



STRIP FOOTING UNDER IN- 
CLINED ECCENTRIC LOAD 



RECTANGULAR FOOTINO OR 
CIRCULAR FOOTING UNDER 
INCLINED ECCENTRIC LOAD 



CIRCULAR FOOTING UNDER 
INCLINED CENTRIC LOAD 



O(0S» / « \* / • %' 

,„= — _ M1.l«N«.,OM,)(l--) .O.. r RN,(l--) 



a„= VERTICAL COMPONENT OF ULTIMATE REARING CAFACITT 

ON EFFECTIVE AR(A FOR ECCENTRIC LOAM 
a * INCLINATION IN OCORCCS OF Q FROM VERTICAL. 
FOR •>« AUUME N T SO 
■ ' « EFFECTIVE FIOTH OF FOOTINO 
L' ■ EFFECTIVE LENGTH OF FOOTINO 



FOR BACKWARD ECCENTRICITY. COMPUTE q„ USING EITHER III NEGATIVE SIGN FOR » 
AND EFFECTIVE FOOTING DIMENSIONS IS' AND L'l OR <2I POSITIVE SION FOR « AND 
ACTUAL FOOTING DIMENSIONS IB AND LI. WHICHEVER GIVES THE LOWER VALUE. 

FOR INCLINED CENTRIC LOADS IN EQUATIONS I AND 2 USE ACTUAL B AND L VALUES 
INSTEAD OF EFFECTIVE VALUES IB' AND L'l 



III 



12) 



RCCTAMOUCAJI FOOTINO. 
•■ ■ ■ - S», 
L'.L-» L 



rift 



COMPUTE ■' AMD L' 
UMNO MA TlOt FROM 
ORAFHMLO* 




L *> ' OIAMETER m 



L' 
2R 




U. S. Army Corps of Engineers 

Figure 6-3. Ultimate bearing capacity of shallow foundations under eccentric inclined loads. 



than the width of the foundation, the factor of safety 
should be applied to the net load (total weight of 
structure minus weight of displaced soil). 

c. Stratified subsoils. Where subsoils are 
variable with depth, the average shear strength within a 
depth below the base equal to the width of the loaded 
area controls the bearing capacity, provided the strength 
at a depth equal to the width of the loaded area or lower 
is not less than one-third the average shear strength of 
the upper layer; otherwise, the bearing capacity is 
governed by the weaker lower layer. For stratified 
cohesive soils, calculate th e ultimate bearing capacity 
from the chart in figure 6-6.~| The bearing pressure on the 
weaker lower layer can be calculated by distributing the 
surface load to the lower layer at an angle of 30 degrees 
to the vertical. 

6-4. Tension forces. Footings subjected to a 
sustained uplift force, Tu, should be designed with a 
minimum factor of safety of 1.5 with respect to weight 
forces resisting pullout expressed as 



where W is the total effective weight of soil and concrete 
located within the prism bounded by vertical lines at the 
base of the footing. Use total unit weights above the 
water table and the buoyant unit weight below. If the 
shear resistance on the vertical sides of the prism 
defined above is considered, a minimum safety factor of 
2 should be used. The lateral earth pressure on the 
vertical sides of the prism should not exceed earth 
pressure at rest and should be considered as active 
earth pressure if the soil is not well compacted. 

6-5. Bearing capacity of rock. For a structure 
founded on rock, adequate exploration is necessary to 
determine the number and extent of defects, such as 
joints, shear zones, and solution features. Estimates of 
the allowa ble bearing pressure can be obtained from 
I table 6-11 Conservative estimates of the allowable 
bearing pressure can be obtained from the following 
expression: 



q a = 0.2q u 



(6-2) 



W 
T u 



1.5 



(6-1) 



6-5 



Allowable bearing pressures for jointe d rock can a lso be 
estimated from RQD values using Itable 6-2.1 Local 
experience should always be ascertained. 



TM 5-818-1 / AFM 88-3, Chap. 7 



h 



PLAN VIEW 
SHOOING 
APPLIED LOADS 
ABOVE GROUND 
SURFACE 



1 

% | 

1 ! 

M u [ 

_J !_ 



SECTION 

THROUGH 

FOOTING 



GROUND-RATER TABLE 
AT BASE OF FOOTING 
? 



■ » o 



GROUND SURFACE 



V 



PLAN VIE* 
SHORING 
EFFECTIVE 
AREA OF 
BASE OF 
FOOTING 




(A) SOIL AND LOADING CONDITIONS 
UNIFORM SANDY SILT: 

y = II2.S LB/CU FT = 0.056 TON CU FT 
flu, = SO LB CU FT - 0.025 TON CU FT 
c = 0.06 TON. 50 FT 
4 = 25" 

IBI COMPUTATION OF NET VERTICAL LOAD 

VERTICAL LOAD ABOVE GROUND SURFACE = 
EFFECTIVE HEIGHT OF SOIL ABOVE BASE OF FOOTING 



DL + NORMAL LL: 

V = 113 TONS VERTICAL 
H B = II TONS LATERAL 
H L : 23 TONS LONGITUDINAL 



V = 113.0 TONS 



16 ' 15 * 61 (112.51 



WEIGHT OF CONCRETE (150 LB CU FTI IN FOOTING IN EXCESS OF DISPLACED SOIL 
I (8 " 15 * 2.51 » 12 * 6 • 3.511 1150 - 112.51 



NET VERTICAL LOAD = Iv = 1600 TONS 



(CI COMPUTATION OF ECCENTRICITY (cl AND INCLINATION (<rl 
TAKING MOMENTS ABOUT 0: 
IM. 11(5*61 



IV 160 

i.M u 23(5 » 61 



IV 160 



IH : "Vim 2 » I23I 2 = 25.5 TONS 



B' = B - 2. B = 8 - 1 52 = 6 48 FT 

L' = L - 2«, = 15 - 3.16 - 11.84 FT 



IH 25 5 

a - ARCTAN = ARCTAN = ARCTAN 0. 159 = 9* 

IV 160 



(D) COMPUTATION OF VERTICAL COMPONENT OF ULTIMATE BEARING CAPACITY 
q = CN |1 * 0.3— | ♦ yON„ [l -— 1 • 0.4yB'N [l --] 

- 06 * 24 [l -t 3 I * (0.056 ■ 6 * 131 I 1 ♦ 10.4 ■ 025 - 6 48 x 10] 

L \ "»/ J V «>/ 

- [I 67 ♦ 4.36] (0 81) * 10 65) (0.411 = 4.88 ♦ 0.27 = 5 15 TONS SQ FT 

IE) COMPUTATION OF FACTOR OF SAFETY WITH RESPECT TO BEARING CAPACITY 

iV 160 



B)' 



ACTUAL BEARING PRESSURE - 



B'L' 6 48 * 11.84 



= 2.09 TONS SO FT 



% 5 15 

FACTOR OF SAFETY = F.S. = — = = 2 S > 2.0 REQUIRED F.S. FOR DL • NORMAL LL 

1- 209 



NOTE: COMPUTATION SHOULD BE REPEATED FOR DL ♦ MAXIMUM LL FS. SHOULD BE GREATER 
THAN 15 FOR THIS CONDITION 



U. S. Army Corps of Engineers 

Figure 6-4. Example of bearing capacity computation for inclined eccentric load on rectangular footing. 



6-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



■$^£ 



vzm 



mom an 



m 



mmm5Z& 



BEARING CAPACITY FACTORS, N FOR FOUNDATIONS IN CLAY (0 = O*) 



3,'L 



1 (SQUARE OR CIRCLE] 

0.5 

(STRIP FOOTING) 



9.0 
8.2 

7.5 



NOTE: BEARING CAPACITY FACTORS BASED ON A SMOOTH BASE AND 
D/B GREATER THAN 4 

B = WIDTH OF FOOTING 
L = LENGTH OF FOOTING 
D = DEPTH OF FOOTING 



0/L = cBN c 1- yOB t 2Dl $ (STRIP LOADINGI 

Q = cB ! N t + yDB 2 + 4BDfj (SQUARE LOADINGI 

Q = cITR 2 N c + yDTTR 2 + 2»RD( (CIRCULAR LOADINGI 



THE SKIN FRICTION, f , IS USUALLY TAKEN AS ONE-HALF 
THE UNCONFINED COMPRESSIVE STRENGTH OF THE CLAY 
FOUNDATION. FOR PILES THE VALUE OF f SHOULD NOT 
EXCEED THE MINIMUM ADHESION VALUE GIVEN IN PARA. 
GRAPH 46b. BECAUSE SKIN FRICTION, f, , IS NOT ALWAYS 
RELIABLE.IT IS OFTEN IGNORED. 



(a) DEEP FOUNDATION IN HOMOGENEOUS CLAY 







■ Off 2 A 










a 






i 






A 






c 


> 






• 




1 


'."..' . ■ . ■ 



%&% it&ziMi&a 



IMW 






&OWW0WIWW 



WEAK OVERBUROEN SOILS 
ISKIN FRICTION NEGLECTEDI 



COMPLETE EMBEDMENT IN SAND 



PARTIAL EMBEDMENT IN SAND 



Q/L = yOBN + O.Sy B 2 N + 2of, (STRIP LOADINGI 0/ L = y DBN + O.Sy B 2 N y + 2hf (STRIP LOADINGI 

Q = yOB 2 N_ + 0.4VB 3 N V + 4BD(. (SQUARE LOADINGI Q = yOB 2 N_ + 0.»yB ! N v + 4Bhf. (SQUARE LOADING] 

/ q i y % i q * y % 

Q = yD«TR 2 N + 0.6yirR 3 N„ + 2irRDf. (CIRCULAR LOADINGI Q = yD*R 2 N„ + 0.6yffR 3 N,, + 2>TRhf.(CI RC UL AR LOADINGI 

/ q i y I ' q ' y \ 

f, = l/2KyD TAN 5 f, = Ky(D - tl/ 2) TAN 5 

WHERE K = COEFFICIENT OF EARTH PRESSURE DEPENDENT ON _ 

DENSITY OF SAND AND METHOD OF FOUNDATION 

PLACEMENT (SEE CHAPTER 12) 
N AND N y - BEARING CAPACITY FACTORS FOR SHALLOW 

FOUNDATIONS (SEE FIGURE 6.1) 

h - ANGLE OF FRICTION BETWEEN SAND AND FOUNDATION iS < <£ ) 



(b) DEEP FOUNDATION IN SAND 



U. S. Army Corps of Engineers 



Figure 6-5. Ultimate bearing capacity of deep foundations. 



6-7 



TM 5-818-1 / AFM 88-3, Chap. 7 



E 

z 






< 
o 

? 

5 

Si! 

o 

k. 

5 

o 

z 




0.2 



O.t 



as o.i i.o 1.2 

LOWER LAYER SHEAR STRENGTH 
UPPER LAYER SHEAR STRENGTH 



Vc, 



10 



u 

E 



O 

u 

2 



u 

< 

u 



z 



o 
O 

Z 

















^ 


S 




6 0S 












■""" 


_ |^ 

H/2R.02 


M 




-^^^^^ 


r 1 MI2R.0.3 I 


*** 


yS 

>/ 


^ 




s^ 








M/2R£0.i— * 






>> 


y 














^r^ .^^ 


^ 












(b) CIRCULAR rOOTINGS 









0.2 



04 



0* 0« 10 12 

LOWER LAYER SHEAR STRENGTH 
UPPER LAYER SHEAR STRENGTH 



It 



It 



'% 



2 



U. S. Army Corps of Engineers 

Figure 6-6. Bearing capacity factors for strip and circular footings on layered foundations in clay. 



6-8 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 6-2. Allowable Bearing Pressure for Jointed Rock 

ROD tsf 

100 300 

90 200 

75 120 
50 65 

25 30 

10 

a If tabulated value exceeds unconfined compressive strength of intact samples of the 
rock, allowable pressure equals unconfined compressive strength. 

U. S. Army Corps of Engineers 

6-9 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 7 
DEWATERING AND GROUNDWATER CONTROL 



7-1. General. The following topics concerning 
foundation design using dewatering and groundwater 
control techniques are discussed in the latest revision 
of TM 5-81 8-5 / NAVFAC P-41 8 / AFM 88-5, Chap. 6: 

a. Excavations requiringdrainage. 

b. Seepage control. 

c. Seepage cutoffs. 



d. Control of surface waters. 

e. Sheet-pile cofferdams. 

f. Foundation underdrainage and 
waterproofing. 

7-2. Foundotion problems. The problems con- 
cerning dewatering or groundwater control should be 
referred to the above-mentioned manual. 



7-1 



TM 5-818-1 / AFM 88-3. Chap. 7 



CHAPTER 8 



SLOPE STABILITY ANALYSIS 



8-1. General. This chapter is concerned with 
characteristics and critical aspects of the stability of 
excavation slopes; methods of designing slopes, 
including field observations and experience, slope 
stability charts, and detailed analyses; factors of safety; 
and methods of stabilizing slopes and slides. The 
emphasis in this chapter is on simple, routine 
procedures. It does not deal with specialized problems, 
such as the stability of excavated slopes during 
earthquakes. 

8-2. Slope stability problems. Excavation slope 
instability may result from failure to control seepage 
forces in and at the toe of the slope, too steep slopes for 
the shear strength of the material being excavated, and 
insufficient shear strength of subgrade soils. Slope 
instability may occur suddenly, as the slope is being 
excavated, or after the slope has been standing for some 
time. Slope stability analyses are useful in sands, silts, 
and normally consolidated and overconsolidated clays, 
but care must be taken to select the correct strength 
parameter. Failure surfaces are shallow in cohesionless 
materials and have an approximately circular or sliding 
wedge shape in clays. 

a. Cohesionless slopes resting on firm soil or 
rock. The stability of slopes consisting of cohesionless 
soils depends on the angle of internal friction §', the 
slope angle, the unit weight of soil, and pore pressures. 
Generally, a slope of 1 vertical (V) on 1 1/2 horizontal (H) 
is adequate; but if the slope is subjected to seepage or 
sudden drawdown, a slope of 1V on 3H.is commonly 
employed. Failure normally occurs by surface raveling or 
shallow sliding. Where consequences of failure may be 
important, required slopes can be determined using 
simple infinite slope analysis. Values of (j)' for stability 
analyses are determined from labor atory tests or 
estimated from correlations ( bara 3-6)71 Pore pressure 
due to seepage reduces slope stability, but static water 
pressure, with the same water level inside and outside 
the slopes, has no effect. Benches, paved ditches, and 
planting on slopes can be used to reduce runoff 
velocities and to retard erosion. Saturated slopes in 
cohesionless materials may be susceptible to 
liquefaction and flow slides during earthquakes, while dry 
slopes are subject to settlement and raveling. Relative 
densities of 75 percent or large r are requir ed to ensure 
seismic stability, as discussed in lChapteHT . 

b. Cohesive slopes resting on firm soil or rock. 
The stability of slopes consisting of cohesive soils 



depends on the strength of soil, its unit weight, the slope 
height, the slope angle, and pore pressures. Failure 
usually occurs by sliding on a deep surface tangent to 
the top of firm materials. For relatively high slopes that 
drain slowly, it may be necessary to analyze the stability 
for three limiting conditions: 

(1) Short-term or end-of-construction 
condition. Analyze this condition using total stress 
methods, with shear strengths determined from Q tests 
on undisturbed specimens. Shear strengths from 
unconfirmed compression tests may be used but 
generally may show more scatter. This case is often the 
only one analyzed for stability of excavated slopes. The 
possibility of progressive failure or large creep 
deformations exists for safety factors less than about 
1.25 to 1.50. 

(2) Long-term condition. If the excavation 
is open for several years, it may be necessary to 
analyze this condition using effective stress methods, 
with strength parameters determined from S tests or R 
tests on undisturbed specimens. Pore pressures are 
governed by seepage conditions and can be determined 
using flow nets or other types of seepage analysis. Both 
internal pore pressures and external water pressures 
should be included in the analyses. This case generally 
does not have to be analyzed. 

(3) Sudden drawdown condition, or other 
conditions where the slope is consolidated under one 
loading condition and is then subjected to a rapid change 
in loading, with insufficient time for drainage. Analyze 
this condition using total stress methods, with shear 
strengths measured in R and S tests. Shear strength 
shall be based on the minimum of the combined R and S 
envelopes. This. case is not normally encountered in 
excavation slope stability. 

c. Effect of soft foundation strata. The critical 
failure mechanism is usually sliding on a deep surface 
tangent to the top of an underlying firm layer. Short-term 
stability is usually more critical than long-term stability. 
The strength of soft clay foundation strata should be 
expressed in terms of total stresses and determined 
using Q triaxial compression tests on undistu rbed 
specimens or other methods described ii j chapter"?] 

8-3. Slopes In soils presenting special problems. 

a. Stiff-fissured clays and shales. The 
shearing resistance of most stiff-fissured clays and 
shales may be 



8-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



far less than suggested by the results of shear tests on 
undisturbed samples. This result is due, in part, to prior 
shearing displacements that are much larger than the 
displacement corresponding to peak strength. Slope 
failures may occur progressively, and over a long period 
of time the shearing resistance may be reduced to the 
residual value-the minimum value that is reached only at 
extremely large shear displacements. Temporary slopes 
in these materials may be stable at angles that are 
steeper than would be consistent with the mobilization of 
only residual shear strength. The use of local 
experience and empirical correlations are the most 
reliable design procedures for these soils. 

b. Loess. Vertical networks of interconnected 
channels formed by decayed plant roots result in a high 
vertical permeability in loess. Water percolating 
downward destroys the weakly cemented bonds between 
particles, causing rapid erosion and slope failure. 
Slopes in loess are frequently more stable when cut 
vertically to prevent infiltration. Benches at intervals can 
be used to reduce the effective slope angle. Horizontal 
surfaces on benches and at the top and bottom of the 
slope must be sloped slightly and paved or planted to 
prevent infiltration. Ponding at the toe of a slope must be 
prevented. Local experience and practice are the best 
guides for spacing benches and for protecting slopes 
against infiltration and erosion. 

c. Residual soils. Depending on rock type and 
climate, residual soils may present special problems with 
respect to slope stability and erosion. Such soils may 
contain pronounced structural features characteristic of 
the parent rock or the weathering process, and their 
characteristics may vary significantly over short 
distances. It may be difficult to determine design shear 
strength parameters from laboratory tests. 
Representative shear strength parameters should be 
determined by back-analyzing slope failures and by using 
empirical design procedures based on local experience. 

d. Highly sensitive clays. Some marine clays 
exhibit dramatic loss of strength when disturbed and can 
actually flow like syrup when completely remolded. 
Because of disturbance during sampling, it may be 
difficult to obtain representative strengths for such soils 
from laboratory tests. Local experience is the best guide 
to the reliability of laboratory shear strength values for 
such clays. 

e. Hydraulic fills. Se e Chapter 15 . 
8-4. Slope stability charts. 

a. Uniform soil, constant shear strength, (j> = 
0, rotational failure. 

(1 ) Groundwater at or below toe of slope. 
Determine shear strength from unconfined compression, 
or better, from Q triaxial co mpression tests. Use the 
upper diagram o fl figure S^TI to compute the safety factor. 
If the center and depth of the critical circle are d esired, 
obtain them from the lower diagrams o fl figure 8^ 

(2) Partial slope submergence, seepage 



surcharge loading, tension cracks. The effect of partial 
submergence of a slope is given by a factor |i w i n] figurel 
|8-2j seepage is given by a factor |i w ' in | figure 8-2| 
surcharge loading is given by a factor ji q in | figure ^2\ 
and tension cracks is given by a factor \i x in | figure 0^ 
Compute safety factor from the following: 

F- |XwUw!_Ua-Ut_No_C (8-1) 

yH + q - y w H w ' 
where 

y = total unit weight of soil 

q = surcharge loading 

N = stability number fror h figure 8-fl 
If any of these conditions are absent, their corresponding 
i factor equals 1 .0; if seepage out of the slope does not 
occur, H. equals IH. 

b. Stratified soil layers, § = O, rotational 
failure. 

(1) Where the slope and foundation 
consist of a number of strata, e ach having a constant 
shear strength, the charts given ir l figures {Til through 8- 
3 can be used by computing an equivalent average 
shear strength for the failure surface. However, a 
knowledge of the location of the failure surface is 
required. The coordinates of the center of the circular 
fai lure surface can be obtained from the lower diagrams 
of Ifioure 8-1.1 The failure surface can be constructed, 
and an average shear strength for the entire failure 
surface can be computed by using the length of arc in 
each stratum or the number of degrees intersected by 
each soil stratum as a weighing factor. 

(2) It may be necessary to calculate the 
safety factor f or failure su rfaces at more than one depth, 
as illustrated ir l figure 8-"4l 

c. Charts for slopes in uniform soils with ty > 0. 

(1) A stability chart for slopes in soils with 
<|> > is shown in | figure 8-5.I Correction factors for 
surcharge loading at the t op of the slop e, submergence, 
and see page are g iven i ri figure iP2\. and for tension 
cracks, ir l figure 8-"3l 

(2) The location of the critical circle can be 
obtained, i f desired, from the plot on the right side of 
figure 8-51 Because simple slopes in uniform soils with § 



> generally have critical circles passing thro ugh the toe 
of the slope, the stability numbers given in l figure 8^ 
were developed by analyzing toe circles. Where subsoil 
conditions are not uniform and there is a weak layer 
beneath the toe of the slope, a circle passing beneath 
the toe may be more critical than a toe circle. 

d. Infinite slopes. Conditions that can be 
analyzed accur ately using charts for infinite slope 
analyses shown In figure 81 6 are- 



8-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



(1) Slopes in cohesionless materials 
where the critical failure mechanism is shallow sliding or 
surface raveling. 

(2) Slopes where a relatively thin layer of 
soil overlies firmer soil or rock and the critical failure 
mechanism is sliding along a plane parallel to the slope, 
at the top of the firm layer. 

e. Shear strength increasing with depth and § 
= 0. A chart for slopes in soils with shear stren gth 
increasing with depth and + = is shown in l figure 8^7] 

8-5. Detailed analyses of slope stability. 

If the simple methods given for estimating slope stability 
do not apply and site conditions and shear strengths 
have been determined, more detailed stability analyses 
may be appropriate. Such methods are described in 
engineering literature, and simplified versions are 
presented below. 

a. The method of moments for ty = 0. This 
method is simple but useful for the analysis of circular 
slip surfaces in <f> = soils, as shown ir l figure 8-"8l 

b. The ordinary method of slices. This simple 
and conservative procedure for circular slip surfaces can 
be used in soils with § >_0. For flat slopes with high pore 
pressures and (j) > 0, the factors of safety calculated by 



this method may be much smaller than values calculated 
by more accurate methods . An example is presented in 
I figures 8-9 Ithrough 8-11.1 Various trial circles must be 
assumed to find the critical one. If § large and c is small, 
it may be desirable to replace the circular sliding surface 
by plane wedges at the active and passive extremities of 
the sliding mass. 

c. The simplified wedge method. This method 
is a simple and conservative procedure for analyzing 
noncircular surfaces. An example is shown in 8-12. 
Various trial failure surfaces with different locations for 
active and passive wedges must be assumed. The base 
of the central sliding wedge is generally at the bottom of 
a soft layer. 

8-6. Stabilization of slopes. If a slide is being 
stabilized by flattening the slope or by using a buttress or 
retaining structure, the shear strength at time of failure 
corresponding to a factor of safety of 1 should be 
calculated. This strength can be used to evaluate the 
safety factor of the slope after stabilization. Metho ds for 
stabilizing slopes and landslides are summarized in ltablel 
8-1 . Often one or more of these schemes may be used 
together. Schemes I through V are listed approximately 
in order of increasing cost. 



8-3 



TM 5-818-1 / AFM 88-3, Chap. 7 




BASE 
CIRCLE 



60 50 40 30 20 
Slope Angle - (deg) 

STABILITY NUMBER 




cot /3 

025 050 l I.O I I.5 

t 1. I II Mi l 



2 346l0d> 
J fl I I I I 




46I0» 
I J I I 

90 80 70 60 50 4Q 30 20 10 

Slope Angle - /3 (d eg) 



9080 70 605040302010 O 
Slope Angle -/3 (c!ag) 

CENTER COORDINATES FOR CRITICAL CIRCLE 

U. S. Army Corps of Engineers 

Figure 8-1. Slope stability charts for§ = soils. 
8-4 



TM 5-818-1 / AFM 88-3, Chap. 7 



I.O 



^09 



£0.8 



REDUCTION FACTORS FOR SURCHARGE LOADING Oz q ) 
£;0> 



Key Skelch 



^ 




J 






i 

:3p' 

i 




^ 






~— | i 














r^ 




.60* 


T^ A r 


irrti 






S^. 


I 















(o) 



OJ 0.2 0.3 0.4 0.5 
Rolio Q/yH 



1.0 
















d 


sO>- 


"7 






1 | 












1.0 
















L , | 




4.0.9 








0.5i 


















.0 




jf0.8 


- B 


ose 


Pirr'« 














L.II; 
















Tirm bose * 

wwrrrr n mrrw-r m tirrrft hi ip hiIh >rww 



00 



0.1 0.2 03 0.4 0.5 
Rolio Vr H 



REDUCTION FACTORS FOR SUBMERGENCE ( Fw ) AND SEEPAGE C/i' w ) 




Key Sketches 



(0 



-£ 10 



0.5 IX) 

Ratios h w/h ond h w/h 




0»oh 



Firm bo« i 



*09 

o 

u 0.8 
•2 









d 


= ffl 


"7 












«^^ 






I 
10 




-^ 


^ 


^ 










0.5 






X 

























D 


ose 


Cir 

1 


















cie~ 



















!H 



Firm boae t 



0»<3H 



0.5 
Rotios H w /h otxJ Hw/h 



LO 



U. S. Army Corps of Engineers 

Figure 8-2. Reduction factors ([i q , ix w , \i w ') for slope stability charts for (j> =0 and§ >0 soils. 



8-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



REDUCTION FACTOR FOR TENSION CRACK 
No Hydrostotic Pressure in Crock 



to 
















Q 


■0' 


7 




















3o; 


-09 








1 - 


v. 08 










_6Q* 


o ° 
o 07 


- 1 
















90* 


i 


o 
""06 


roe 


Circle 














0"> 




1 


1 


i 















Key Sketch 



(o) 



O.I 0.2 0.3 0.4 0.5 
Rotio h »/h 



to 


















d-co-j 


0.9 
^■0.8 
?Q7 

^■0.6 

0.5 


















-1.0 


















05 





















u 




P 


ose 

1 


Ci 

1 


















1 
















/^Tension CfOCks 



X. » .. fWf? R? . *,* 



(t>) 



0.1 0.2 0.3 0.4 0.5> 

Ratio H t/H 



1.0 

£09 
fc0.8 
I 0.7 

0.5 



(0 



REDUCTION FACTOR FOR TENSION CRACK 
Full Hydrostolic Pressure in Crock 

0*0% 



















30* 


r^ 


















60° 




















<V5* 




— Tc 


e C 


*ircl 





























Key Sketch 



OJ 0.2 03 0.4 0.5 
Rotio h */h 



Tension cocks 



1.0 


















d'cp-j 


-0.9 
%0 


















_Ui_ 


















0.5 


o 

«07 

°"0.6 
05 


















^"cT" 


B 


ose 


c„ 


cle 































h y> 


A<m 


X K 1 |H, 




* 


WW 


~~s&- 






D«dH 


Firm bOie 



0.1 0.2 03 0.4 0.5 
(d) Rotio H(/ H 

U. S. Army Corps of Engineers 

Figure 8-3. Reduction factors (tension cracks, nJ for slope stability charts for§ = and§ > O soils. 



8-6 



TM 5-818-1 / AFM 88-3, Chap. 7 




120 LB/ FT 
C u = BOO PSF 



CRITICAL CIRCLE 
FOUND USING CHARTS 



y - 100 LB/ FT 3 
400 PSF 



CIRCLE USED FOR 
AVERAGING STRENGTHS 



C aup = O4)(600) t (16) (400) ± (66X500) . 49g psf 

.3 



Y 



"ave 96 

120 + 100 + 105 



ave 



108 lb/ft J 



12 



d = 5f=0.5 H w /H=2f=0.33 



From figure 8-2, v = 0.95 for d = 0.5 
and H w /H = 0.33 w 

From figure 8-1, N Q = 5.6 for d = 0.5 and 6 = 50° 



F - (0-95)(5.6)(498) . , „ 
h (108)(24) - (62.4)(8) " '•" 



From figure 8-1, x Q = 0.35 , y Q = 1.5 , critical 
circle intersects near toe of slope 

x Q = (0.35)(24) = 8.4 ft, y Q = (1.5)(24) = 36 ft 



CRITICAL CIRCLE 



y = I 20 LB/ FT 
C = 600 PSF 




CIRCLE USED FOR 
AVERAGING STRENGTHS 



C aye = (22)(600H ( 62 H*°°) = 452 psf 



r_* : ^ J ^= HO lb/ft 3 
8 



'ave 2 
d = H, ,/H 



w 



24 



0.33 



From figure 8-2, u = 0.93 for 6 = 50° 
and H w /H =0.33 w 

From figure 8-1, N = 5.8 for e = 50° 

f - (0.98)(5.8)(452) . .. 
F - (110)(24) - (62.4)(8) = 1J4 

(more critical than circle 
tangent to elev. -20 ft) 

From figure 8-3, x Q = 0.35 , y Q = 1.4 , critical circle 
intersects near toe of slope 

x = (0.35H24) - 8.4 ft, y Q = (1.4)(Z4) = 33.6 ft 



U. S. Army Corps of Engineers 

Figure 8-4. Example of use of charts forslopes in soils with uniform strength and§ = 0. 



8-7 



TM 5-818-1 / AFM 88-3, Chap. 7 



300 
200 



100 



50 — 



m 
< 

h 

« 10 

_l 
< 



FOR c = 0: 1 










- P 1 ^r^' 








- 


F = — - btan 


*>*"^ 










~ / S^ 












~/y^^' 










- 


^^ 


















x = P e tan^ 




C0 c 








1 





H 



— 



SLOPE RATIO b = cot /3 



-t 



-<: 

O 



rrrr 



(IN FORMULA 



H' 



7H+q-7 w H w 
^qM„M t 

7H+ Q-7 W H^ 




2 3 

SLOPE RATIO b 



CENTER COORDINATES FOR CRITICAL CIRCLE 



FOR P„ TAKE q = 0, U = I FOR UNCONSOLIDATED CONDITION) 
e q 



U. S. Army Corps of Engineers 



Figure 8-5. Slope stabdilty charts forty > soils. 



8-8 



TM 5-818-1 / AFM 88-3, Chap. 7 




»* Cysundroined strength 
V 



Steps: 

^ Extropolote strength profile upward to determine volue of Hq, 
where strength profile intersects zero 

(§) Colcolote M * HqM 

® Determine stability number N from choM below 

(4) Determine C b « strength at bottom of slope 

® Colcolote F - N j;^ 



USe y * y buOyonl f0f submer 9 ed *'°P e 
Use Y*Y m lor no water outside slope 
Use overoge y for portly submerged slope 




60 30 

/3(deg) 



U. S. Army Corps of Engineers 



Figure 8-6. Stability charts for infinite slopes. 



8-9 



TM 5-818-1 / AFM 88-3. Chap. 7 




y - totol unit weight of soil 
y w * unit weight of water 

c'« cohesion intercept •»_,. 
, v lEffective 

9 E friction angle J Stress 

r u * pore pressure ratio * ~ 
u s pore pressure at depth H 



Steps: 

(J) Determine r u from measured 
pore pressures or formulas 
at right 

(D Determine A ond B from 
charts below 

(D Colculote F « A 1^4' + B-^r 



Seepoge porollel to slope 






K ** aJ 




Seepage emerging from slope 
at angle 9 

Xw 1 



u y 1 + tan /3 tan d 



1.0 

09 

08 

07 

< 06 

* 05 

01 

04 
03 
02 
Ql 
O 



Tjj'O 



E 
o 









0| 












n ? 












_0 3_ 












■ Ci A- 






\ 


/> 




-r\ «^ - 






//, 


'/l 




U.D 








// 


<*■ 


^06: 






ill 


7> 


S*~ 












/ 












11 


7 











12 3 4 5 

Slope Ratio b= cot Q 



o 














9 
8 

co7 

«- 6 




































^ 














§4 














a 3 
2 


\ 












\ 












i 






























12 3 4 5 

Slope Rotio b - cot /3 



U. S. Army Corps of Engineers 

Figure 8- 7. Slope stability charts for§ = and strength increasing with depth. 



8-10 



TM 5-818-1 / AFM 88-3. Chap. 7 



C u = 600psf 




Cu* 400 psf 



-20-J 



Section AreoUt 2 } y (lb/ft 3 ) Weight (lb/ft) Moment Arm(ft) Moment (f I -to/ft) 



<§> 


444 


120 


53,280 


+ 33 


+ L76M0 6 


® 


456 


100 


45,600 


+ 23 


+ I.05M0 6 


© 


564 


105 


59,220 





0.0 


© 


356 


62.4 


20,970 


-19 


-0.40*I0 6 



Totol Overturning Moment « + 2.4UI0 6 



Moment 
Section Ave.Length(ft) Cu(psf) Force(lb/ft) A rm« Radius (ft) Moment (ft-lb/ft) 



® 


14 


600 


8,400 




60 


0.50*I0 6 


© 


16.5 


400 


6,600 




60 


0.40M0 6 


© 


69 


500 


34,500 




60 


2.07 » I0 6 


© 


18 







Totol 


Resist! 


60 


0.00 




rig Moment ■ 


2.97«I0 6 




Fnefnr nf 5i 


ifalw. F » 


Resisting 


Moment 


2.97»l0 6 , 


.23 



Overturning Moment 2.4UI0 6 



U. S. Army Corps of Engineers 



Figure 8-8. Method of moments forty = 0. 



8-11 



TM 5-818-1 / AFM 88-3, Chap. 1 



Loyer y (lb/ft 3 ) c (lb/ft 2 ) <f> (degrees) 




in -io 



-2 CM 



U. S. Army Corps of Engineers 



Figure 8-9. Example problem for ordinary method of slices. 



8-12 



TM 5-818-1 / AFM 88-3, Chap. 7 



, b 

• m 



LAYER 



' h. 



y. = unit weight of layer i 

h. = height of layer at center of slice 



W. = partial weight = bh.Y. 
i li 



EW. = total weight of slice 



Slice 


b 


hi 




Yi 


Wi 


EWi 


No. 


ft 


ft 




lb/ft3 


lb/ft 


lb/ ft 


1 


15 


5 




110 


8,100 


8,200 


2 


15 


13 




110 


21,400 


21,400 


3 


15 


4 




105 


6,300 








17. 


5 


110 


28,900 


35,200 


4 


15 


11. 


5 


105 


18,100 








19. 


5 


110 


32,200 


50,300 


5 


15 


4 

15 




110 
105 


6,600 
23,600 








19. 


5 


110 


32,200 


62,400 


6 


15 


11. 
15 


5 


110 
105 


19,000 
23,600 








17. 


5 


110 


28,900 


71,500 


7 


15 


15 
15 




110 
105 


24,800 
23,600 








15 




110 


21,400 


69,800 


8 


15 


15 
15 




110 
105 


24,800 
23,600 








5 




110 


8,200 


56,600 


9 


16 


15 




110 


26,400 








7. 


5 


105 


12,600 


39,000 


10 


11 


7. 


5 


110 


9,100 


9,100 



U. S. Army Corps of Engineers 

Figure 8-10. Example of use of tabular form for computing weights of slices. 



8-13 



TM 5-818-1 / AFM 88-3, Chap. 7 























-s- 


























c 
u 
























3 


3 


















u 






O 


O 


















Zero 


for 




to 


0) 






Slice 


W 


I 


a 


C 


* 


Total Stress 


W cos a 





u 






No. 


kip/ft 
8.2 


ft 
17.5 


deg 
-32 


kip/ft 2 
0.75 


deg 
5 


Analy 


sis 


Ml kip/ ft 
6.9 


3 


3 


cJl 
13.12 


W sin a 


1 


0.61 


-4.35 


2 


21.4 


16.2 


-22 


0.75 


5 






19.8 




1.73 


12.15 


-8.22 


3 


35.2 


15.5 


-13 


0.75 


5 






34.3 




3.00 


11.62 


-7.92 


4 


50.3 


15.1 


-4 


0.75 


5 






50.2 




4.39 


11.32 


-3.51 


5 


62.4 


15.1 


4 


0.75 


5 






62.2 




5.44 


11.32 


4.35 


6 


71.5 


15.5 


13 


0.75 


5 






69.7 




6.10 


11.62 


16.08 


7 


69.8 


16.2 


22 


0.75 


5 






64.7 




5.66 


12.15 


26.15 


8 


56.6 


17.5 


32 


0.75 


5 






48.0 




4.20 


13.12 


30.00 


9 


39.0 


22.0 


43 


0.10 


30 






28.5 




16.45 


2.20 


26.60 


10 


9.1 


18.5 


55 


0.06 


35 






5.2 




3.64 


1.11 


7.45 



I 51.2 99.7 86.8 






V^ 



c » cohesion intercept* 

<|> » friction angle \ at base of 

slice 

u = pore pressure 



p _ Z(W cos a - uil) tan <)> + £c& 
ZW sin a 



F -l|g4-1.74 



86.8 



Figure 8-11. Example of use of tabular form for calculating factor of safety by ordinary method of slices. 



8-14 



TM 5-818-1 / AFM 88-3, Chap. 7 



ranee omgrmb 



p»»vt toot 



CENTRAL BLOCK 



ACTIVE *f OGC 




* -to.«oo tan »* i -f.no LB 
«. - » K - U, CM /«r - yjl TAN * ♦ * «,M 4 TAW (or ♦ y 1 

-ai»o.*a>-».taocosnl t«nw> t aw mm tm it 

■ 12.100 LB 



0; »0 

N, • |B, - U,> TAN *, » t, L., 

- obi ,soo- 110.1001 tan r» 





■ oi.ooo ta» or * 01.000 lb 
b, - 1 Jo, - u, w* (at - ^|r«# *, * i «, h, ta« («r - ^-) 

• l 111 JOB - ttJM IM Ml TAB BT ♦ I OHO) II TAN II" 



o, ■■, tab fur •-pi 

* >«.!«» TAM4T* 14,100 LB 
B, i 1«, H, TAMJ 

■ 2 MOW (111 TA« «T « 12.000 LB 



-f-V) 



F ACT CM O* lAFtTV 




U 4 ■ •.100 LB 
n 4 , w ■ OH LB/OQ 



MOTBT: OBIVWQ ANO BjSlSTBtO) FOBCtt COMFVTtO 

USB* SATURATES BCKBfT OB 10IL M BCOOtl 
AMO CENTRAL BLOCK ANO TOTAL URLBT 



ABALTOIO BMOBN FOR CONDITION OF STEAOV SEEPAOE 
U*M0 A SBOFLlFlCO ANALYSIS BASED ON THE AMUUB- 
TKB» THAT tOLNBOTCNTnu. LBHH ABC VEBTICAL.. 
AMALYSJI SHORN FOB FAILURE ALONO PLANE] A-O-C-O. 
ANALYSE! MUST BE REPEATED FOB OTmCB FOTEnTiAl 
FAILURE FLAMES TO OCTCBNONC CRITICAL SURFACE 
BMCM BCSJLT9 Bt BBIBBJM FACTOB OF SAFETY. 



Figure 8- 12. Example of simplified wedge analysis. 



8-15 



TM 5-818-1 / AFM 88-3, Chap. 7 



I. Excavation 



ARpMcjbl* Methods 



1. Reduce slope height by excavation at top 

of slope. 

2. Flatten the slope angle. 

3. Excavate a bench in upper part of slope, 
U. Excavate the entire slide mass. 



Area has to be accessible to construction 
equipment. Disposal site needed for exca- 
vated soil- Drainage sometimes Incorporated 
in this method. 



II. Drainage 



f 



1. Small diameter, horizontal drains 
{hydraugers) . 



2. Continuous deep subdraln trench. 
Generally 3 to 15 ft deep. 



3. Drilled vertical wells - generally IB- to 
36-lrwdlaaeter. 



Improve surface drainage along top of 
slope with open ditch or paved gutter. 
Install deep-rooted, erosion- resistant 
plants on slope face. 



1. Host effective if can tap natural aquifer. 

Drains are usually free-flowing. 

2. Trench bottom should be sloped to drain 

and be tapped with an outlet pipe. Per- 
forated pipe should be placed on trench 
bottom. Top of trench should be capped 
with impervious material. 

3. Can be pumped or tapped with a gravity 

outlet. Several wells In a row, joined 
at bottom can form a drainage gallery. 
Top of each well should be capped with 
Impervious material. 

6, Good practice for moat slopes. Direct 
the discharge away from slide mass. 



Ill . Earth or rock 
buttress (or 
berm fill) 




Excavate slide mass and replace with 
compacted earth or rock buttress fill. 
Toe of buttress must be keyed into 
firm soil or rock below slide plane. 
Drain blanket with gravity flow outlet 
is provided in back slope of buttress 
fill. 

Conpacted earth or rock berm placed at 
and beyond the toe. Drainage may be 
provided behind berm. 



Access for construction equipment and 
temporary stockpile area required. 
Excavated soil can usually be used in 
fill, Underpinning of existing structures 
may be required. Might have to be done In 
shore sections if stability during con- 
struction is critical. 

Sufficient width and thickness of berm 

required so failure will not occur below 
' or through berm. 



IV. Retaining structures 



^T 




Retaining wall - crib or cantilever type* 



Drilled, cast-in-place vertical piles, 
bottomed well below bottom of slide 
plane. Cenerally 18 to 36 In. in 
diameter and 4- to 8-ft spacing. 



Drilled, cast-in-place vertical piles 

tied back with battered piles or a dead- 
man. Piles bottomed well below slide 
plane. Cenerally 12 to 30 in. in 
diameter and at 4- to 8— ft spacing. 

Earth anchors and rock bolts. 



Reinforced earth. 



Usually expensive. Cantilever walls 
might have to be tied back. 

Spacing should be such that soil can arch 
between piles. Grade beam can be used to 
tie piles together. Very large diameter 
(6 ft +) piles have been used for deep 
slides. 

Space close enough so soil will arch 

between piles. Piles can be tied together 
with grade beam. 



U. Can be used for high slopes, and in very 

limited stress. Conservative design should 

be used, especially for permanent support. 



5. Usually expensive. 



Special techn iques 



1. Grouting 

2. Chemical Injection 

3. Electroomosis (In fine-grained soils). 

4. Freezing 

5. Heating 



I. and 2. Used successfully in a number of 

cases. Used at other times wl th 
little success. At the present, 
theory Is not completely understood. 

3. Generally expensive. 

4. and b. Special methods which must be 

specifically evaluated at each site. 
Can be expensive. 

All of these techniques should be carefully 

evaluated in advance to determine the probable 
cost and effectiveness. 



(Courtesy of W. J. Turnbull and M . J. Hvorslev, "Special Problems in Slope Stability, " Journal, Soil 
Mechanics and Foundation Division, Vol93, No. SM4, 1 967, pp 499-528. Reprinted by permission of 

the American Society of Civil Engineers, New York.) 



Table 8-1. Methods of Stabilizing Slopes and Landslides 

8-16 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 9 



SELECTION OF FOUNDATION TYPE 



9-1. Foundation - selection considerations. 

Selection of an appropriate foundation depends upon the 
structure function, soil and groundwater conditions, 
construction schedules, construction economy, value of 
basement area, and other factors. On the basis of 
preliminary information concerning the purpose of the 
structure, foundation loads, and subsurface soil 
conditions, evaluate alternative types of foundations for 
the bearing capacity and total and differential 
settlements. Some foundation alternatives 



for different subsoil conditions are summarized in ltablel 
9-1. 

a. Some foundation alternatives may not be 
initially obvious. For example, preliminary plans may not 
provide for a basement, but when cost studies show that 
a basement permits a floating foundation that reduces 
consolidation settlements at little or no increase in 
construction cost, or even at a cost reduction, the value 
of a basement may be substantial. Benefits of basement 
areas include needed garage space, office or stor- 



Table 9-1. Foundation Possibilities for Different Subsoil Conditions 







Foundation Poss 


ihilifies 


Subsoil Conditions 


Light, Flexible Structure 


1. 


Heavy, Rigid Structure 


Deep compact or 


Footing foundations 


Footing foundations 


stiff deposit 






2. 


Shallow mat 


Deep compressible 


1. 


Footing foundations 


1. 


Deep mat with possi- 


strata 




on compacted granular 
zone 3 




ble rigid construc- 
tion in basement 




2. 


Shallow mat 3 


2. 


Long piles or cais- 




3. 


Friction piles 


3. 


sions to by-pass 
Friction piles 


Soft or loose 


1. 


Bearing piles or 


1. 


Bearing piles or 


strata overly- 




piers 




piers 


ing firm strata 


2. 
3. 


Footing foundations 
on compacted granular 
zone 3 
Shallow mat 2 


2. 


Deep mat 


Compact or stiff 


1. 


Footing foundations 3 


1. 


Deep mat (floating 


layer overlyinga 








type) 


soft deposit 


2. 


Shallow mat 3 


2 


Long piles or cais- 
sons to by-pass soft 
deposit 


Alternating soft 


1. 


Footing foundations 3 


1. 


Deep mat 


and stiff 


2. 


Shallow mat 3 


2. 


Piles or caissons to 


layers 








underlying firm 



stratum to provide 

satisfactory 

foundation 



3 Consider possible advantages of site preloading, with and without vertical sand drains to accelerate 
consolidation. 

(Courtesy of L. J. Goodman and R. H. Karol. Theory and Practice of Foundation Engineering , 
1968, p 312. Reprinted by permission of Macmillan Company, Inc New York.) 



9-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



age space, and space for air conditioning and other 
equipment. The last item otherwise may require 
valuable building space or disfigure a roofline. 

b. While mat foundations are more expensive 
to design than individual spread footings, they usually 
result in considerable cost reduction, provided the total 
area of spread footings is a large percentage of the 
basement area. Mat foundations may decrease the 
required excavation area, compared with spread 
footings. 

c. The most promising foundation types 
should be designed, in a preliminary manner, for detailed 
cost comparisons. Carry these designs far enough to 
determine the approximate size of footings, length and 
number of piles required, etc. Estimate the magnitude of 
differential and total foundation movements and the 
effect on structure. The behavior of similar foundation 
types in the area should be ascertained. 

d. Final foundation design should not be 
started until alternative types have been evaluated. Also, 
the effect of subsurface conditions (bearing capacity and 
settlement) on each alternative should be at least 
qualitatively evaluated. 

e. A checklist of factors that could infl uence 
foundation selection for family housing is shown in ltablel 
9-2. 

9-2. Adverse subsurface conditions. If poor soil 
conditions are encountered, procedures that may be 
used to ensure satisfactory foundation performance 
include the following: 

a. Bypass the poor soil by means of deep 
foundations extending to or into a suitable bearing 
material (chap. 11). 

b. Design the structure foundations to 
accommodate expected differential settlements. 
Distinguish between settlements during construction that 
affect a structure and those that occur during 
construction before a structure is affected by differential 
settlements. 

c. Remove the poor material, and either treat 
and replace it or substitute good compacted fill material. 



d. Treat the soil in place prior to construction 
to improve its properties. This procedure generally 
requires considerable time. The latter two procedures 
are carried out usin g various techniques of soil 
stabilization described in lchapterTfe . 

9-3. Cost estimates and final selection. 

a. On the basis of tentative designs, the cost 
of each promising alternative should be estimated. 
Estimate sheets should show orderly entries of items, 
dimensions, quantities, unit material and labor costs, and 
cost extensions. Use local labor and material costs. 

b. The preliminary foundation designs that are 
compared must be sufficiently completed to include all 
relevant aspects. For example, the increased cost of 
piling may be partially offset by pile caps that are smaller 
and less costly than spread footings. Similarly, mat or 
pile foundations may require less excavation. 
Foundation dewatering during construction may be a 
large item that is significantly different for some 
foundation alternatives. 

c. The most appropriate type of foundation 
generally represents a compromise between 
performance, construction cost, design cost, and time. 
Of these, design cost is generally the least important and 
should not be permitted to be a controlling factor. If a 
lower construction cost can be achieved by an alternative 
that is more expensive to design, construction cost 
should generally govern. 

d. Foundation soils pretreatment by 
precompression under temporary surcharge fill, 
regardless of whether vertical sand drains are provided 
to accelerate consolidation, requires a surcharge loading 
period of about 6 months to a year. The time required 
may not be available unless early planning studies 
recognized the possible foundation cost reduction that 
may be achieved. Precompression is frequently 
advantageous for warehouses and one-story structures. 
Precompression design should be covered as a separate 
design feature and not considered inherent in structure 
design. 



9-2 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 9-2. Checklist for Influence of Site Charateristics on Foundation Selection for Family Housing 



Foundations 



Site Characteristics 



Natural Ground 

Level 

Rolling 
Rolling 

Hilly 
Hilly 

Groundwater 
Surface 



Footing level 
below foot- 
ing level 

Soil Type 

GW, GP, GM, GC 
SW, SP, SM, SC 

ML, CL, OL 
MH, CH, OH 



Grading 

None 

None 
Cut and fill 

None 
Cut and fill 



Post 



1, 2 



Spread 



1, 2, 3, 4, 5 
1, 2, 3, 4, 5 



Requires temporary 
lowering 



1, 2 



3, 4, 5, 6 3, 4, 5, 6 



Slab-on-Grade (all) 



Requires grading 
1, 2, 3, 4, 5 

Requires grading 
1, 2, 3, 4, 5 



1, 2 

3, 4, 5, 6 



Basement 



1, 2, 3, 4, 5 

1, 2, 3, 4, 5 

1, 2, 3, 4, 5 

1, 2, 3, 4, 5 

1, 2, 3, 4, 5 



Do not use 



Use perimeter 
drainage 



1, 2 

3, 4, 5, 6 



1. Compaction control - increase density if required, use compaction control in fills. 

2. Check relative density of cohesionless (GW, GP, SW, SP) soils; generally based on standard 
penetration resistance. 

3. Use undrained shear strength, s , to estimate bearing capacity and stress ratios for slab 
design. 

4. Check if settlement is a problem. 

5. Check liquidity index as indication of normally or preconsolidated clay. 

6. Check expansive properties. 



U. S. Army Corps of Engineers 



9-3 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 10 



SPREAD FOOTINGS AND MAT FOUNDATIONS 



10-1. General. When required footings cover more 
than half the area beneath a structure, it is often 
desirable to enlarge and combine the footings to cover 
the entire area. This type of foundation is called a raft or 
mat foundation and may be cheaper than individual 
footings because of reduced forming costs and simpler 
excavation procedures. A mat foundation also may be 
used to resist hydrostatic pressures or to bridge over 
small, soft spots in the soil, provided the mat is 
adequately reinforced. Although mat foundations are 
more difficult and more costly to design than individual 
spread footings, they can be used effectively. 

10-2. Adequate foundation depth. The foundation 
should be placed below the frost line (chap 18) because 
of volume changes that occur during freezing and 
thawing, and also below a depth where seasonal volume 
changes occur. The minimum depth below which 
seasonal volume changes do not occur is usually 4 feet, 
but it varies with location. If foundation soils consist of 
swelling clays, the depth may be considerably greater, 
as described in TM 5-818-7. On sloping ground, the 
foundation should be placed at a depth such that it will 
not be affected by erosion. 

10-3. Footing design. 

a. Allowable bearing pressures. Procedures 
for determini ng allowable bearing pressures are 
presented in Ichapter 6~~| In many instances, the 
allowable bearing pressure will be governed by the 
allowable settlement. Criter ia for determ ining allowable 
settlement are discussed in I chapter~5~] The maximum 
bearing pressure causing settlement consists of dead 
load plus normal live load for clays, and dead load plus 
maximum live loads for sands. Subsoil profiles should 
be examined carefully to determine soil strata 
contributing to settlement. 

b. Footings on cohesive soils. 

(1) If most of the settlement is anticipated 
to occur in strata beneath the footings to a depth equal to 
the distance between footings, a settlement analysis 
should be made assuming the footings are independent 
of each other. Compute settlements for the maximum 
bearing pressure and for les ser values. A n example of 
such an analysis is shown in l figure 10-11 If significant 
settlements can occur in strata below a depth equal to 
the distance between footings, the settlement analysis 
should consider all footings to determine the settlement 
at selected footings. Determine the vertical stresses 



beneath individual footings from the influence charts 
presented in Ichapter 5~l The footing size should be 
selected on the basis of the maximum bearing pressure 
as a first trial. Depending on the nature of soil 
conditions, it may or may not be possible to proportion 
footings to equalize settlements. The possibility of 
reducing differential settlements by proportioning footing 
areas can be determined only on the basis of successive 
settlement analyses. If the differential settlements 
between footings are excessive, change the layout of 
the foundation, employ a mat foundation, or use piles. 

(2) If foundation soils are nonuniform in a 
horizontal direction, the settlement analysis should be 
made for the largest footing, assuming that it will be 
founded on the most unfavorable soils disclosed by the 
borings and for the smallest adjacent footing. Structural 
design is facilitated i f results o f settlement analyses are 
presented in charts Kfig 10-1) which relate settlement, 
footing size, bearing pressures, and column loads. 
Proper footing sizes can be readily determined from such 
charts when the allowable settlement is known. After a 
footing size has been selected, compute the factor of 
safety with respect to bearing capacity for dead load plus 
maximum live load condition. 

c. Footings on cohesionless soils. The 
settlement of footings on cohesionless soils is generally 
small and will take place mostly during construction. A 
procedure for proportioning footings on sands to restrict 
the differential settlemen t to within tolerable limits for 
most structures is given in | figure IQ"^ 

d. Foundation pressures. Assume a planar 
distribution of foundation pressure for the structural 
analysis of a footing. This assumption is generally 
conservative. For eccentrically loaded footings, the 
distribution of the bearing pressure should be determined 
by equating the downward load to the total upward 
bearing pressure and equating the moments of these 
forces about the center line in accordance with 
requirements of static equilibrium. Examples of the 
b earing press ure distribution beneath footings are shown 
in lfigure 10751 

10-4. Mat foundations. 

a. Stability. The bearing pressure on mat 
foundations should be selected to provide a factor of 
safety of 



10-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



at least 2.0 for dead load plus normal live load and 1.5 
for dead load plus maximum live load. By lowering the 
base elevation of the mat, the pressure that can be 
exerted safely by the building is correspondingly 
increased (chap 11), and the net increase in loading is 
reduced. The bearing pressure should be selected so 
that the settlement of the mat foundation will be within 
limits that the structure can safely tolerate as a flexible 
structure. If settlements beneath the mat foundation are 



medium '£JP 



CLAY — «-0* 

r . i te lb/cu fTc. a«o t/sq ft 



Ah 



. O-JS. «..1.10 



MARINO CAFAOTY ANALYSIS 

«. . l.*eNf • |ON< (SQUARE FOOTMOI 

Jt.» . 0.40 > «-^«^JA ■ T « l)- S.» T/JO FT 
WAX SAFEK«fUNOP*ESSlSME>* a >4 ( /F9 
4 S -Wf - I.EJ T/M FT FOR DL . NORMAL U. 
4, -A* -MI T/SQ FT FOR OL. MAX U. 



more than the rigidity of the structure will permit, a 
redistribution of loads takes place that will change the 
pressure distribution beneath the structure, as 
subsequently described. The bearing capacity of loose 
sands, saturated silts, and low-density loess can be 
altered significantly as a result of saturation, vibrations, 
or shock. Therefore, the allowable bearing pressure and 
settlement of these soils cannot be determined in the 
usual manner for the foundation soils 



settlement analysis 

neolecting settlement of sano layer, 
compute influence value («« | for stress 
at mo-font of clay stratum ifio. ib. 

wothof so ■/• 

footino - ft m4n ». w. 



StTTLEMEN 



0.011 
0.0* 
8.101 
0.14S 



0.072 
O.MO 

0.SS4 

R. ♦ 4R.A. 



EQUATION 1 



IT.AK-— '- m !•,„-* 2-* 

F. -< T "' ,< ^ 1)»"« T, > - OLRM T/SO FT 
O JO M* » 4*. t. 



4, "77TTo ,,,0 " ta ». 







"\*/* 


^ 










► 






\- \» 


V 






MAX SAFE 
MARINE 

RMRWRR 


^/\ 


^ 







30 



! 

I IS 

3 



10 



1 , 


y 


/ 




M« MMMH 1 


c »•*•*••-*. 


/ 

/ 










t 


i»2Ji>' 






/ 










/ 
/ 




" I m. . 















I 4 • • 

WS9TN OF FOOTMO IN FT 



10 



10 



EXAMPLE 0* CHARTS FOR SELECTING ALLOWABLE BEARING PRESSURES AND FOOTING SIZES 
RESULTWC Ml EQUAL SETTLEMENT OF FOOTINGS. CURVES BASED ON EQ I. 



■OTE: •armoo for comstructhm oesmw charts apflicaoui only msRE most confriisiru 

STRATA ARE AOOVE IONS WHERE ITMUa KhEATH FOOTimOS OVCAUkF 

AFTER FOOTMO MIES ARC SELECTED ON THE BASIS OF SCTTLCMNT. FOOTIMM WNHILO BE 
CMECKEO TO ASSURE THAT THE COLUMN LOAD IOL • MAX LU OaVKXU OY THE FOOTIMO AREA 
ODES MOT EXCEED l» T/SO FT 



U. S. Army Corps of Engineers 

Figure 10-1. Example of method for selecting allowable bearing pressure. 

10-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



may be subject to such effects. Repl ace or stab ilize 
such foundation soils, as discussed ii j chapter~H 6, if 
these effects are anticipated. 

b. Conventional analysis. Where the 
differential settlement between columns will be small, 
design the mat as reinforced concrete flat slab assuming 
planar soil pressure distribution. The method is generally 
applicable where columns are more or less equally 
spaced. For analysis, the mat is divided into mutually 
perpendicular strips. 

c. Approximate plate analysis. When the 
column loads differ appreciably or the columns are 
irregularly spaced, the conventional method of analysis 
becomes seriously in error. For these cases, use an 



analysis based on the theory for beams or plates on 
elastic foundations. Determine the subgrade modulus by 
the use of plate load tests. The method is suitable, 
particularly for mats on coarse-grained soils where 
rigidity increases with depth. 

d. Analysis of mats on compressible soils. If 
the mat is founded on compressible soils, determination 
of the distribution of the foundation pressures beneath 
the mat is complex. The distribution of foundation 
pressures varies with time and depends on the 
construction sequence and procedure, elastic and 
plastic deformation properties of the foundation concrete, 
and 



fo) 0,/6-t 



*. 







ir-so 


/ 


| 




/, 


ti-40 


' | 






I 
n'to 


r i 




// 


Y | 
N-20 




// 


N'tS 


N-IO 






ft-S 




I 



ft) 0,/fO.S 





1 


tfSO 


1 






H-40 




1 


/tt-30 




A 


/n'20 




// 








' lv-5 




1 



12 3 4 
Width- of footing, t, ft 





(el t 


w* 


'0.2S 












ti-30 


1 




1 




/ 


f 


/ 


"1 

N-40 


1 




/ 


J 


/ 


N-30 




( 


/, 




J 


1 


VsU 




tJ-13 


^ 


N-IO 


k 


*\ 






N-S 






1 



o . Design chut for proportioning shallow footing! on sand. 



if., 
Correction factor C M " Tff 



04 0.6 at 1.0 1.2 1.4 16 1.1 tj) 



z 


OS 




1.0 


1. 

i 1 


IS 


2.0 
2.S 
3.0 


i 


3.5 


i 


4 




4.5 




5.0 



^-— -~ 


z 


7 


t^ 

j 


7 


t 




1 


I 


I 



1. Determine H values at 2-1/2-ft in- 
tervals between base of footing and 
depth E below base. Calculate 
average N value. 

2. Select allowable soil pressure from 
design chart (a) based on settlement 
of 1 in. 

3- If effective overburden pressure cor- 
responding to depth of footing differs 
greatly from 1 ton/sq ft, adjust N 
value according to chart (b). 

k. Multiply allowable soil pressure by 
correction factor for depth to water 



table- 



C = 

v 



0.5 + 0.5 



D, + B 



° ■ Chart for correction of A'-valurs in 
sand for influence of overburden pressure 



(Courtesy of R. B. Peck, W. E. Hanson, and T. H. Thornburn, 
Foundation Engineering , 1974, p 312. Reprinted by permis- 
sion of John Wiley & Sons, Inc., New York.) 



Figure 10-2. Proportioning footings on cohesionless soils. 



10-3 



TM 5-818-1 / AFM 88-3, Chap. 7 



£ w ft 




P, -P,-2- 




(1) e» . « y . P - CONSTANT 



P 2 



12) e<i- 



P.-Efj-St) 
P 2 .*(1*SS) 




(2) e,< L 



P..P e = i(i,&L) 




/"" FO< 



NO PRESSURE BETWEEN 
FOOTING AND SOIL 



~ UUU1 W1^ 



(3) e -i. 

6 



P, -0 



p, -ii 



Pj' 



*-S(ns) 



DISTRIBUTION OF BEARING PRESSURES BENEATH STRIP FOOTINGS 



NO PRESSURE BETWEEN FOOTING 
AND SOIL 




6 
e > B 



NOTE: NO ZERO PRESSURE WILL EXIST BENEATH THE 
FOOTING IF THE LOAD. W. IS APPLIED WITHIN 
THE RHOMBUS SHOWN AT RIGHT 



A PART OF THE 
SOIL PRES5URE 
WILL BE ZERO 
(SEE REF 20 FOR 
COMPUTATION 
OF PRESSURES) 



LjTHli, 

T 6 

DISTRIBUTION OF BEARING PRESSURES BENEATH RECTANGULAR FOOTINGS 



Figure 10-3. Distribution of bearing pressures. 



10-4 



TM 5-818-1 / AFM 88-3, Chap.7 



time-settlement characteristics of foundation soils. As a 
conservative approach, mats founded on compressible 
soils should be designed for two limiting conditions: 
assuming a uniform distribution of soil pressure, and 
assuming a pressure that varies linearly from a minimum 
of zero at the middle to twice the uniform pressure at the 
edge. The mat should be designed structurally for 
whichever distribution leads to the more severe 
conditions. 

10-5. Special requirements for mat foundations. 

a. Control of groundwater. Exclude 
groundwater from the excavation by means of cutoffs, 
and provide for temporary or permanent pressure relief 
and dewatering by deep wells or wellpoints as described 
in TM 5-818-5/AFM 88-5, ( Dhapter 6~l Specify 
piezometers to measure drawdown levels during 
construction. Specify the pumping capacity to achieve 
required drawdown during various stages of 
construction, including removal of the temporary system 
at the completion of construction. Consider effects of 
drawdown on adjoining structures. 

b. Downdrag. Placement of backfill against 
basement walls or deep raft foundations constructed in 
open excavations results in downdrag forces if weight of 
backfill is significant with respect to structural loading. 
Estimate the downdrag force on the basis of data in 

I chapter TH . 

10-6. Modulus of subgrade reaction for footings 
and mats. 

a. The modulus of su bgrade re action can be 
determined from a plate load test Kpara 4^61) using a 1- by 
1 - foot plate. 



ksiB 



(10-1) 



where 
ksf 



<sl 



the modulus of subgrade reaction for 
the prototype footing of width B 
the value of the 1 - by 1 -foot plate in 
the plate load test 

The equation above is valid for clays and assumes no 

increase in the modulus with depth, which is incorrect, 

and may give k1 , which is too large. 

For footings or mats on sand: 



( B + 1 
V 2B 



(10-2) 



For a rectangular footing or mat of dimensions of B x 
mB: 

( 15m5)(10-3) 

with a limiting value of k sf = 0.667k S |. 
b. k s may be computed as 

k s = 36q a . (kips per square foot) 
(10-4) 



which has been found to give values about as reliable as 
any method. This equation assumes q a (kips per square 
foot) for a settlement of about 1 inch with a safety factor, 
F = 3. A typical range of values of k s is given in table 3- 
7. 

10-7. Foundations for radar towers. 

a. General. This design procedure provides 
minimum footing dimensions complying with criteria for 
tilting rotations resulting from operational wind loads. 
Design of the footing for static load and survival wind 
load conditions will comply with other appropriate 
sections of this manual. 

b. Design procedure. This design procedure 
is based upon an effective modulus of elasticity of the 
foundation. The effective modulus of elasticity is 
determined by field plate load tests as described in 
subparagraph d below. The design procedure also 
requires seismic tests to determine the S-wave velocity 
in a zone beneath the footing at least 1 1/2 times the 
maximum size footing required. Field tests on existing 
radar towers have shown that the foundation performs 
nearly elastically when movements are small. The 
required size of either a square or a round footing to 
resist a specific angle of tilt, a, is determined by the 
following: 



B J = 4320(F) M 



/ 1 - M 2 \ (square 
^ E s ' footing) 



D 3 = 6034(F) M / 1 -M 



M / I - IVI \ 

a I E s ) 



(round 
footing) 



(10-5) 
(10-6) 



where 
B, D 



size and diameter of footing, 

respectively, feet 
F = factor of safety (generally use 2.0) 
M = applied moment at base of footing 

about axis of rotation, foot-pounds 
a = allowable angle of tilt about axis of 

rotation, angular mils (1 angular mil = 

0.001 radian) 
E s = effective modulus of elasticity of 

foundation soil, pounds per cubic foot 
The design using equations (10-5) and (10-6) is only 
valid if the seismic wave velocity increases with depth., 
If the velocity measurements decrease with depth, 
special foundation design criteria will be required. The 
discussion of these criteria is beyond the scope of this 
manual. 

c. Effective modulus of elasticity of foundation 
soil (E s ). Experience has shown that the design modulus 
of elasticity of in-place soil ranges from 1000 to 500, kips 
per square foot. Values less than 1000 kips per square 
foot will ordinarily present severe settlement problems 
and are not satisfactory sites for radar towers. Values in 
excess of 5000 kips per square foot may be encountered 
in dense gravel or rock, but such values are not used in 
design. 



10-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



(1) Use equations (10-5) and (10-6) to 
compute- 

(a) Minimum and maximum footing sizes 
using E s = 1000 and 5000 kips per square foot, 
respectively. 

(b) Two intermediate footing sizes using 
values intermediate between 1000 and 5000 kips per 
square foot. 

Use these four values of B or D in the following 
equations to compute the increase (or pressure change) 
in the live load, AL. 

square footing AL = 17.0M (pounds per 



B J 



square foot) (10-7) 



round footing AL = 20.3M (pounds per 

D 3 square foot) (10-8) 



(2) The E s value depends on the depth of the 
footing below grade, the average dead load pressure on 
the soil, and the maximum pressure change in the live 
load, AL, on the foundation due to wind moments. A 
determination of the E, value will be made at the 
proposed footing depth for each footing size computed. 

(3) The dead load pressure, q , is computed as 
the weight, W, of the radar tower, appurtenances, and 
the footing divided by the footing area, A. 



qo 



EW 
A 



(10-9) 



The selection of loadings for the field plate load test will 
be based on qo and AL. 

d. Field plate load test procedure. The 
following plate load test will be performed at the elevation 
of the bottom of the footing, and the tes t apparatus will 
be as described in TM 5-824-3/AFM 88-6, |ChapteF3l 

(1) Apply a unit loading to the plate equal 
to the smallest unit load due to the dead load pressure 
q . This unit loading will represent the largest size 
footing selected above. 

(2) Allow essentially full consolidation 
under the dead load pressure increment. Deformation 
readings will be taken intermittently during and at the end 
of the consolidation period. 

(3) After consolidation under the dead 
load pressure, perform repetitive load test using the live 
load pressure AL computed by the formulas in paragraph 



10-7c. The repetitive loading will consist of the dead 
load pressure, with the live load increment applied for 1 
minute. Then release the live load increment and allow 
to rebound at the dead pressure for 1 minute. This 
procedure constitutes one cycle of live load pressure 
application. Deformation readings will be taken at three 
points: at the start, after the live load is applied for 1 
minute, and after the plate rebounds under the dead load 
pressure for 1 minute. Live load applications will be 
repeated for 15 cycles. 

(4) Increase the dead load pressure, q , to 
the second lowest value, allow to consolidate, and then 
apply the respective live load increment repetitively for 15 
cycles. 

(5) Repeat step 4 for the remaining two 
dead load pressure increments. 

(6) An uncorrected modulus of elasticity 
value is computed for each increment of dead and live 
load pressure as follows: 

E s ' = 25.5 _AL (1 -\i 2 ) 
S 

E s ' = uncorrected effective modulus of 
elasticity for the loading condition used, pounds per 
square foot 

S = average edge deformation of the plate 
for the applied load, determined from the slope of the last 
five rebound increments in the repetitive load test, inches 

H = Poisson's ratio (se e table 3-6b - 

(7) The above-computed uncorrected 
modulus of elasticity will be corrected for be nding of the 
plate as described in TM 5-824-3/AFM 88-6. IChapteT3l 
where E' is defined above, and E, is the effective 
modulus of elasticity for the test conditions. 

e. Selection of required footing size. The 
required footing size to meet the allowable rotation 
criteria will be determined as follows: 

(1) Plot on log-log paper the minimum and 
the maximum footing size and the two intermediate 
footing sizes versus the required (four assumed values) 
effective modulus of elasticity for each footing size. 

(2) Plot the measured effective modulus of 
elasticity versus the footing size corresponding to the 
loading condition used for each test on the same chart 
as above. 

(3) These two plots will intersect. The 
footing size indicated by their intersection is the minimum 
footing size that will resist the specified angle of tilt. 



10-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 11 



DEEP FOUNDATIONS INCLUDING DRILLED PIERS 



11-1. General. A deep foundation derives its support 
from competent strata at significant depths below the 
surface or, alternatively, has a depth to diameter ratio 
greater than 4. A deep foundation is used in lieu of a 
shallow foundation when adequate bearing capacity or 
tolerable settlements cannot be obtained with a shallow 
foundation. The term deep foundation includes piles, 
piers, or caissons, as well as footings or mats set into a 
deep excavation. This chapter discusses problems of 
placing footings and mats in deep excavations and 
design of drilled piers. Drilled piers (or caissons) are 
simply large-diameter piles, but the design process is 
somewhat different. An arbitrary distinction between a 
pile and pier is that the caisson is 30 inches or more in 
diameter. 

11-2. Floating foundations. A foundation set into a 
deep excavation is said to be compensated or floating if 
the building load is significantly offset by the load of soil 
removed during excavation. The foundation is fully 
compensated if the structural load equals the load 
removed by excavation, partially compensated if the 
structural load is greater, and overcompensated if the 
structural load is less than the weight of the excavated 
soil. A compensated foundation requires a study of 
expected subsoil rebound and settlement, excavation 
support systems, means to maintain foundation subsoil 
or rock integrity during excavation, and allowable bearing 
pressures for the soil or rock. 

11-3. Settlements of compensated foundations. 

a. The sequence of subsoil heave during 
excavation and subseque nt settlemen t of a deep 
foundation is illustrated in Ifigure 11-"Kh ). If effective 
stresses do not change in the subsoils upon the initial 
excavation, i.e., the soil does not swell due to an 
increase in water content, and if no plastic flow occurs, 
then only immediate or elastic rebound from change in 
stress occurs. If the structural load is fully compensated, 
the measured settlement of the foundation would consist 
only of recompression of the elastic rebound, generally a 
small quantity, provided subsoils are not disturbed by 
excavation. 

b. If the negative excess pore pressures set 
up during excavation "dissipate, " i.e., approach static 
values, before sufficient structural load is applied, 
foundation swell occurs in addition to elastic rebound. 
(The original effective stresses will decrease.) The 



foundation load recompresses the soil, and settlement of 
the foundation consists of elastic a nd consolidation 
components as shown in | figure 11 -Kb ). Consolidation 
occurs along the recompression curve until the 
preconsolidation stress is reached, whereupon it 
proceeds along the virgin compression curve. Calculate 
the foundation heave and subse quent settlement using 
procedures outlined in | chapter ~5\ 

c. If the depth with respect to the type and 
shear strength of the soil is such that plastic flow occurs, 
loss of ground may develop around the outside of the 
excavation with possible settlement damage to 
structures, roads, and underground utilities. 

d. The rate and amount of heave may be 
estimated from the results of one-dimensional 
consolidation tests; however, field evidence shows that 
the rate of heave is usually faster than predicted. A 
study of 43 building sites found that the field heave 
amounted to about one-third the computed heave. 
Where excavations are large and are open for 
substantial time before significant foundation loadings 
are applied, th e actual he ave may be close to the 
computed heav $. Figure 1U 2 is a plot of a series of field 
results of heave versus excavation depth, in which the 
heave increases sharply with the depth of excavation. 
An example of heave and subsequent settlement 
calculation s for a compensated foundation is shown in 

Ifigure 5-4] 

e. The yielding of the excavation bottom can 
be .caused by high artesian water pressures under the 
excavation or by a bearing capacity failure resulting from 
the overburden pressure on the soil outside the 
excavation at subgrade elevation. Artesian pressure can 
be relieved by cutoffs and dewatering of the underlying 
aquifer using deep wells. The pumped water may be put 
back in the aquifer using recharge wells outside the 
excavation perimeter to avoid perimeter settlements or to 
preserve the groundwater table for environmental 
reasons, but this operation is not simple and should be 
done only when necessary. 

f. The likelihood of bearing capacity failure 
exists pri marily in cla yey soils and should be analyzed as 
shown in lchapterTft . A factor of safety, F s >2, should 
ideally be obtained to minimize yielding and possible 
settlement problems. A large plastic flow may cause the 
bottom of the excavation to move upwards with re- 



11-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



suiting loss of ground. To avoid this possibility, 
investigate- 

(1) Potential for plastic flow, i.e., 
relationship between shear stress and shear strength. 

(2) Sequence of placing wall bracing. 

(3) Depth of penetration of sheeting below 
base of excavation. 

g. Two commonly used procedures to control 
bottom heave are dewatering and sequential excavation 
of the final 5 feet or more of soil. Groundwater lowering 
increases effective stresses and may reduce heave. 
Where subsoil permeabilities are not large, a deep and 
economical lowering of the groundwater to minimize 



heave can sometimes be achieved by an educator-type 
wellpoint system. Permitting a controlled rise of the 
groundwater level as the building loan is applied acts to 
reduce effective stresses and counteracts the effect of 
the added building load. Sequential excavation is 
accomplished by removing soil to final grade via a series 
of successive trenches. As each trench is opened, the 
foundation element is poured before any adjacent trench 
is opened. This procedure recognizes the fact that more 
heave occurs in the later excavation stages than in 
earlier stages and is frequently used in shales. 

h. The tilting of a compensated foundation can 
occur if structural loads are not symmetrical or if soil 



£ 

> 
o 



Excavation 



Building Construction 




Time 



Elastic Settlement 



(a) No effective stress change during excavation 




Consolidation 
Settlement 



Total 
Settlement 



Elastic Settlement 

i 



Time 

(b) Decrease in effective stress during excavation 
(causes subsoil rebound and leads to additional 
settlement) 

Note: Figures (a) and (b) are illustrative only. 
Some rebound may continue after start of 
building construction until sufficient 
building load has been applied. Additional 
settlement exceeding that shown in figures 
occurs if building load exceeds weight of 
excavation. 

U. S. Army Corps of Engineers 

Figure 11-1. Effect of pore pressure dissipation during excavation and settlement response. 



11-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



conditions are nonuniform. Tilting can be estimated from 
settlement calculations for different locations of the 
excavation. Control of tilt is not generally necessary but 
can be provided by piles or piers, if required. Bearing 
capacity is not usually important unless the building is 
partially compensated and founded on clay. The factor 
of safety against bearing failure is calculated (see chap 6 
for quit) and compared with the final total soil stress 
using the building load, qo, less the excavation stress 
as follows: 



FS 



q u n 

qo-yD 



(11-1) 



The factor of safety should be between 2.5 and 3.0 for 
dead load plus normal live load. 

/'. Settlement adjacent to excavations 
depends on the soil type and the excavation support 
system method employed (chap 14). With properly 
installed strutted or anchored excavations in 



cohesionless soils, settlement will generally be less than 
0.5 percent of excavation depth. Loss of ground due to 
uncontrolled seepage or densification of loose 
cohesionless soils will result in larger settlements. 
Surface settlements adjacent to open cuts in soft to firm 
clay will occur because of lateral yiel ding and mov ement 
of soil beneath the bottom of the cut. Figure H^ can be 
used to estimate the magnitude and extent of settlement. 

11-4. Underpinning. 

a. Structures supported by shallow 
foundations or short piles may have to be underpinned if 
located near an excavati on. Techn iques for 

underpinning are depicted in figure 11-4~1 The most 
widely accepted methods are jacked down piles or piers, 
which have the advantage of forming positive contact 
with the building foundation since both can be 
prestressed. The use of drilled piers is of more recent 
vintage and is more economical where it can be used. In 
sandy soils, chemical 



2.5 



h 
li. 

Q 

Z 
D 

o 

CD 

llf 

a. 

z 
o 

(- 
< 
> 
< 
u 

X 

w 



FINE-GRAINED SOILS 
SHALES 




200 



U. S. Army Corps of Engineers 



Figure 1 1-2. Excavation rebound versus excavation depth. 



11-3 



TM 5-818-1 / AFM 88-3. Chap. 7 



Z 

UJ 
UJ 

t— 
I— 

UJ 







z 
o 
^ 1 

< 
> 
< 

u 
x 

UJ 

u_ 2 
O 




12 3 

DISTANCE FROM EXCAVATION 
DEPTH OF EXCAVATION 



4 
/o 



ZONE I - SAND AND SOFT TO HARD 
; CLAY, (s u > 500 LB/SQ FT) 

ZONE U - VERY SOFT TO SOFT CLAY 
(s u < 500 LB/SQ FT) 

LIMITED DEPTH OF C LAY 
BELOW BASE OF EXCAVATION 
SIGNIFICANT DEPTH OF CLAY 
BELOW BASE OF EXCAVATION 
WHERE Z±L < 5 



1) 
2) 



ZONE HI - VERY SOFT TO SOFT CLAY, 

500 LB/SQ FT) 
SIGNIFICANT DEPTH OF CLAY 



(s u < 
1) - 



BELOW BASE OF 
AND WHERE 2^ 



EXCAVATION 
> 5 



U. S. Army Corps of Engineers 



Figure 1 1-3. Probable settlements adjacent to open cuts. 



11-4 



TM 5-818-1 / AFM 88-3, Chap. 7 



injection stabilization (chap 16) may be used to underpin 
structures by forming a zone of hardened soil to support 
the foundation. 

b. In carrying out an underpinning operation, 
important points to observe include the following: 

(1 ) Pits opened under the building must be 
as small as possible, and survey monitoring of the 
building must be carried out in the areas of each pit to 
determine if damaging movements are occurring. 

(2) Care must be taken to prevent 
significant lifting of local areas of the building during 
jacking. 

(3) Concrete in piers must be allowed to 
set before any loading is applied. 

(4) Chemically stabilized sands must not 
be subject to creep under constant load. 

c. The decision to underpin is a difficult one 
because it is hard to estimate how much settlement a 



building can actually undergo before being damaged. 
The values given in tables 5-2 and 5-3 may be used as 
guidelines. 

11-5. Excavation protection. During foundation 
construction, it is important that excavation subsoils be 
protected against deterioration as a result of exposure to 
the elements and heavy equipment. Difficulties can 
occur as a result of slaking, swelling, and piping of the 
excavation soils. Also, special classes of soils can 
collapse upon wetting (chap 3). Metho ds for prote cting 
an excavation are described in detail in ltable H^TI alona 
with procedures for identifying problem soils. If these 
measures are not carried out, soils likely will be subject 
to a loss of integrity and subsequent foundation 
performance will be impaired. 




JACKED DOWN PILE 




CHEMICALLY 

STABILIZED 

SOIL 



a. JACKED DOWN PILES 
(PLACED IN SEGMENTS) 



CHEMICAL SOLIDIFICATION 
OF SANDS 





J* 



z 
o 

< 

> 
< 
O 
X 

111 



est 



b. JACKED DOWN PIERS (DRILLED 
AND CAST IN PLACE OR INSTALLED 
BY HAND EXCAVATION) 



d. CAST-IN-PLACE CONCRETE 

SLURRY WALL IN LIEU OF 
CONVENTIONAL UNDERPINNING 



U. S. Army Corps of Engineers 



Figure 11-4. Methods of underpinning. 



11-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 11-1. Excavation Protection. 



Soil Type 


Identification 


Problems and Mechanism 
(1) One dimensional 


(1) 


Preventative Measures 


Overconsolidated 


(1) Stiff plastic clays, 


Rapid collection of sur- 


clays 


natural water content 




heave, maximum 




face water, or grading 


near plastic limit 


heave at excava- 




around excavation 








te) Seefiaure3-14for 




tion center 


(2) 


Deep pressure relief to 




swelling potential 


(2) Swelling dependent 




minimize rebound 








on plasticity 


(3) 


Place 4" - 6" working mat 






(3) 


Usually fractured 
and fissured. Ex- 
cavation opens 
these, causing 
softening and 
strength loss 


(4) 


of lean concrete immedi- 
ately after exposing sub- 
grade (mat may be placed 
over underseepage and 
pressure relief systems 
placed in sand blanket) 
If sloped walls, use as- 
phalt sealer on vertical 
walls, burlap with rubber 
sheeting or other mem- 
brane on flatter slopes 


Chemically 


(1) Lab testing 


(1) 


Swelling, slaking 


(1) 


Protect from wetting and 


inert, unce- 


(2) Soil and geologic maps 




and strength loss 




drying by limiting area 


mented clay- 


(3) Local experience 




if water 




open at subgrade 


stone or shale 






infiltration 


(2) 


Concrete working mat 






(2) 


Rebound if over- 
consolidated 
(generally the 
case) 


(3) 


Paving, impervious mate- 
rials to avoid water in- 
filtration; place sealing 
coats immediately after 






(3) 


Cracking if dry 




exposure 








or evaporation 


(4) 


Reduce evaporation and 








allowed 




drying to prevent 
swelling on resaturation 






(Continued) 







(Sheet 1 of 3) 



11-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 11-1. Excavation Protection-Continued 



Soil Type 




Identification 


Problems and Mechanism 




Preventative Measures 


Limestone 


(1) 


Geologic maps and 


Open cavities and cav- 


(1) 


Avoid infiltration; col- 






local experience 


erns; soft infilling 




lect surface runoff and 




(2) 


Borings and soundings 


in joints, cavities. 




convey it to a point 






to determine cavity 


Dewatering can cause 




where its infiltration 






locations 


cavity collapse or 
settlements of in- 




will not affect 
excavation 








filling materials 


(2) 
(3) 


Eliminate leaks from util- 
ity or industrial piping 
and provide for inspec- 
tion to avoid infiltra- 
tion continuing after 
correction 

Avoid pumping which 
causes downward seepage 




(1) 




(1) When infiltrated 


(1) 


or recharging 


Collapsing soils 


Ste figure 3-161 


Drainage and collection 


(primarily 




Y d vs W L plot 


with water, sudden 




system to avoid water 


Ibess and vol- 


(2) 


Y d < 85 PCF 


decrease in bulk 




infiltration 


canic ash) 


(3) 


Large open "structure" 


volume; may or may 


(2) 


Preload excavation area, 






with temporary source 


not require load- 




and pond with water to 






of strength 


ing in conjunction 




cause collapse prior to 




(4) 


Liquid limit < 45% 
plasticity index 0-25% 


with seepage 
(2) Water action re- 




excavation 




(5) 


Natural water content 
well below 100% 
saturation 


duces temporary 
source of 
strength, usually 








(6) 


Optimum water content 
for collapse between 
13-39%(3) 


capillary tension 
or root structure 
Low erosion 
resistance 

(Continued) 







(Sheet 2 of 3) 



11-7 



Table 11-1. Excavation Protection-Continued. 



TM 5-818-1 / AFM 88-3, Chap. 7 



Soil Type 




Identification 


Problems and Mecf" 


Soft, normally 


(1) 


Recently deposited, or 


(1) Primarily low 


consolidated 




no geologic loading 


strength, unable 


or sensitive 




and unloading 


to support const. 


clays 


(2) 


Leached marine clays 


equipment 




(3) 


Natural water content 


(2) Remolded by con 






near or above liquid 


struction activ- 






limit 


ity, causing 
strength loss 



Preventative Measures 

Provide support, and prevent 
remolding: place timber 
beams under heavy 
equipment; cover 
excavation bottom with 1 ' 
2' of sand and gravel fill 
and/or 6" - 8" of lean 
concrete 



(Sheet 3 of 3) 



11-8 



TM 5-818-1 / AFM 88-3, Chap. 7 



11-6. Drilled piers. Drilled piers (also drilled shafts or 
drilled caissons) are often more economical than piles 
where equipment capable of rapid drilling is readily 
available, because of the large capacity of a pier as 
compared with a pile. 

a. Pier dimension and capacities. Drilled piers 
can support large axial loads, up to 4, 000 kips or more, 
although typical design loads are on the order of 600 to 
1, 000 kips. In addition, drilled piers are used under 
lightly loaded structures where subsoils might cause 
building heaving. Shaft diameters for high-capacity piers 
are available as follows: 

From 2- 1/2 feet by 6-inch increments 
From 5 feet by 1 -foot increments 

Also available are 15- and 2-foot-diameter shafts. 
Commonly, the maximum diameter of drilled piers is 
under 10 feet with a 3- to 5-foot diameter very common. 
Drilled piers can be belled to a maximum bell size of 
three times the shaft diameter. The bells may be 
hemispherical or sloped. Drilled piers can be formed to a 
maximum depth of about 200 feet. Low capacity drilled 
piers may have shafts only 12 to 18 inches in diameter 
and may not be underreamed. 

b. Installation. The drilled pier is constructed 
by drilling the hole to the desired depth, belling if 
increased bearing capacity or uplift resistance is 
required, placing necessary reinforcement, and filling the 
cavity with concrete as soon as possible after the hole is 
drilled. The quantity of concrete should be measured to 
ensure that the hole has been completely filled. 
Reinforcement may not be necessary for vertical loads; 
however, it will always be required if the pier carries 
lateral loads. A minimum number of dowels will be 
required for unreinforced piers to tie the superstructure to 
the pier. Reinforcement should be used only if 
necessary since it is a construction obstruction. 
Consideration should be given to an increased shaft 
diameter or higher strength concrete in lieu of 
reinforcement. In caving soils and depending on local 
experience, the shaft is advanced by: 

(1) Drilling a somewhat oversize hole and 
advancing the casing with shaft advance. Casing may 
be used to prevent groundwater from entering the shaft. 
When drilling and underreaming is completed, the 
reinforcing steel is placed, and concrete is placed 
immediately. The casing may be left in place or 
withdrawn while simultaneously maintaining a head of 
concrete. If the casing is withdrawn, the potential exists 
for voids to be formed in the concrete, and special 
attention should be given to the volume of concrete 
poured. 

(2) Use of drilling mud to maintain the shaft 
cavity. Drilling mud may be used also to prevent water 
from entering the shaft by maintaining a positive head 
differential in the shaft, since the drilling fluid has a 



higher density than water. The reinforcing steel can be 
placed in the slurry-filled hole. Place concrete by tremie. 
(3) Use of drilling mud and casing. The 
shaft is drilled using drilling mud, the casing is placed, 
and the drilling mud is bailed. Core barrels and other 
special drilling tools are available to socket the pier shaft 
into bedrock. With a good operator and a drill in good 
shape, it is possible to place 30- to 36-inch cores into 
solid rock at a rate of 2 to 3 feet per hour. Underreams 
are either hemispherical or 30- or 45-degree bell slopes. 
Underreaming is possible only in cohesive soils such that 
the underslope can stand without casing support, as no 
practical means currently exists to case the bell. 

c. Estimating the load capacity of a drilled pier. 
Estimate the ultimate capacity, Q, of a drilled pier as 
follows: 



Q u = Q 



+ Qup(point) (1 1"2) 



us(skin resistance) + "°<up(point) 

The design load based on an estimated 1-inch 
settlement is: 



Qd — vJus + wnq 

3 



(11-3) 



11-9 



(1) Drilled piers in cohesive soil. The skin 
resistance can be computed from the following: 

H 

Qus =0tavg I C C z dz 



where 

a avg = factor frorh table jJJ2l 
H = shaft length 
C = shaft circumference 

Cz = undrained shear strength at depth z 

Us e 1 table 11-2 for the length of shaft to be considered in 
computing H and for limiting values of side shear. The 
base resistance can be computed from the following: 
Qu P = N c CbAb (11-5) 

where 

Nc = bearing capacity factor of 9 (table 1 1 - 

2) 
Cb = undrained shear strength for distance 

of two diameters below tip 
A B = base area 

(2) Drilled piers in sand. Compute skin 
resistance from the following: 

H 
Qus = otavg C I P z tan dz (11 -6) 


where 

°=avg = 0.7, for shaft lengths less than 25 

feet 
°=avg = 0.6, for shaft lengths between 25 and 

40 feet 
°=avg = 0.5, for shaft lengths more than 40 
feet 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 1 1-2. Design Parameters for Drilled Piers in Clay 



D. 



Design Category 

Straight-sided shafts in either homo- 
geneous or layered soil with no soil 
of exceptional stiffness below the 
base 

1 . Shafts installed dry or by the 
slurry displacement method 

2. Shafts installed with drilling 
mud along some portion of the 
hole with possible mud entrapment 

Belled shafts in either homogeneous 
or layered clays with no soil of 
exceptional stiffness below the base 

1 . Shafts installed dry or by the 
slurry displacement methods 

2. Shafts installed with drilling 
mud along some portion of the 
hole with possible mud entrapment 

Straight-sided shafts with basedry 
resting on soil significantly 
stiffer than soil around stem 
Belled shafts with base resting 
on soil significantly stiffer than 
soil around stem 



"avg 



Side Resistance 

L i m i t on s i de 

shear - tsf 



0.6 



0.3 



0.3 



(a) 



0.15 



(b) 



2.0 



0.5 1 ' 



0.5 



0.3 



(b) 



Tip 
Resistance 



Remarks 



(a) a avg may be in- 
creased to 0.6 and side 
shear increased to 2.0 
tsf for segments drilled 
dry 



(b) aavg may be in- 
creased to 0.3 and side 
shear increased to 0.5 
tsf for segments drilled 



Note: In calculating load capacity, exclude (1) top 5 ft of drilled shaft, (2) periphery of bell, and (3) bottom 5 ft of 
straight shaft and bottom 5 ft of stem of shaft above bell. 



U. S. Army Corps of Engineers 



11-10 



Q 



up ■ 



A B qt = 1 .31 Bq, (11-7) 
0.6B 



11-11 



TM 5-818-1 / AFM 88-3, Chap. 7 



p z = effective overburden pressure at depth 
z 

(j> = effective angle of internal friction 
Arching develops at the base of piers in sand similar to 
piers in clay; thus, the bottom 5 feet of shaft should not 
be included in the integration limits of the above 
equations. The base resistance for a settlement of about 
1 inch can be computed from the following: 



where 

A B 

B 

qt 
q t 

q t 



base area 
base diameter 
for loose sand 

32,000 pounds per square foot for 
medium sand 
80,000 pounds per square foot for 
dense sand 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 12 
PILE FOUNDATIONS 



12-1. General. Bearing piles are deep foundations 
used to transmit foundation loads to rock or soil layers 
having adequate bearing capacity to support the 
structure and to preclude settlement resulting from 
consolidation of soil above these layers. When the 
bearing strata are below the groundwater table, and 
when offshore structures are being built, piles may be 
the most economical type of deep foundation available 



because they do not require dewatering of the site. Piles 
also may be used to compact cohesionless soils and to 
serve as anchorages against lateral thrust and vertical 
uplift. 

12-2. Design. The selection, design, and placement 
of pile foundations are discussed in detail in the latest 
revision of TM 5-809-7. 



12-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 13 
FOUNDATIONS ON EXPANSIVE SOILS 



13-1. General. Natural and man-made deposits of 
soils that contain substantial proportions of clay minerals 
have a potential for swelling or shrinking with change in 
water content. Certain engineering aspects such as site 
studies, heave predictions, and foundation types are 
discussed in the latest revision of TM 5-818-7. 



13-2. Foundation problems. The problems related to 
expansive soils should be referred to the above- 
mentioned manual. 



13-1 



TM 5-818-1 / AFM 88-3, Chap. 7 
CHAPTER 14 
RETAINING WALLS AND EXCAVATION SUPPORT SYSTEMS 



14-1. Design considerations for retaining walls. 

a. General. Retaining walls must be designed 
so that foundation pressures do not exceed allowable 
bearing pressures, wall settlements are tolerable, safety 
factors against sliding and overturning are adequate, and 
the wall possesses adequate structural strength. 
Methods for evaluating earth pressures on retaining walls 
and design procedures are summarized herein for 
cohesionless backfill materials, which should be used 
whenever practicable. 

b. Forces acting on retaining walls. Forces 
include earth pressures, seepage and uplift pressures, 
surcharge loads, and weight of the wall. Typical load 
diagrams for principal wall types are shown in figure 14- 
1. The magnitude and distribution of active and passive 
earth pressures are developed from the earth theory for 
walls over 20 feet high and from semiempirical curves for 
lower walls. The subgrade reaction along the base is 
assumed linearly distributed. 

14-2. Earth pressures. 

a. Earth pressure at rest. For cohesionless 
soils, with a horizontal surface, determine the coefficient 
of earth pressure at rest, K>, from the following: 



K 0= 1 - sin (j) 



(14-1) 



b. Active earth pressure. Formulas for 
calculating the coefficient of active earth pressure for a 
c ohesionless soil with planar boundaries are presented 
in lfiqure 14-21 

c. Passive earth pressure. Formulas for 
calculating the coefficient of passive earth pressure for a 
c ohesionless soil with planar boundaries are presented 
in lfigure 14-31 

d. Earth pressure charts. Earth pressure 
coefficients based on planar sliding surfaces are 
presented in lfigure 14-4.1 The assumption of a planar 
sliding surface is sufficiently accurate for the majority of 
practical problems. A logarithmic spiral failure surface 
should be assumed when passive earth pressure is 
calculated and the angle of wall friction, 5, exceeds cj>'/3. 
Earth pressure coefficients based on a logarithmic spiral 
sliding surface are presented in textbooks on 
geotechnical engineering. Passive pressure should not 
be based on Coulomb theory since it overestimates 
passive resistance. Because small movements mobilize 
5 and concrete walls are relatively rough, the wall friction 
can be considered when estimating earth pressures. In 
general, values of 5 for active earth pressures should not 



exceed (j>'/2 and for passive earth pressures should not 
exceed (j>'/3. The angle of wall friction for walls subjected 
to vibration should be assumed to be zero. 

e. Distribution of earth pressure. A 
presentation of detailed analyses is beyond the scope of 
this manual. However, it is sufficiently accurate to 
assume the following locations of the earth pressure 
resultant: 

(1) For walls on rock: 

0.38H above base for horizontal or 
downward sloping backfill 
0.45H above base for upward sloping 
backfill 

(2) For walls on soil: 

0.33H above base of horizontal backfill 
0.38H above base of upward sloping 
backfill 
Water pressures are handled separately. 

f. Surcharge loads. Equation s for 
concentrated point and line load are presented ir l figurel 
14-5. For uniform or nonuniform surcharge pressure 
acting on an irregular area, use influence charts based 
on the Boussinesq equations for horizontal loads and 
double the horizontal pressures obtained. 

g. Dynamic loads. The effects of dynamic 
loading on earth pressures are beyond the scope of this 
manual. Refer to geotechnical engineering textbooks 
dealing with the subject. 

14-3. Equivalent fluid pressures. The equivalent 
fluid method is recommended for retaining walls less 
than 20 feet high. Assign availab le backfill material to a 
category listed in Ifigure 14-6. I If the wall must be 
designed without knowledge of backfill properties, 
estimate backfill pressures on the basis of the most 
unsuitable material th at may be us ed. Equivalent fluid 
pressures are shown irTfiqure 14-61 for the straight slope 
backfill and in lfigure 1 4-7 I for the broken slope backfill. 
Dead load surcharges are included as an equivalent 
weight of backfill. If the wall rests on a compressible 
foundation and moves downward with respect to the 
backfill, pressures should be increased 50 percent for 
backfill types 1 , 2, 3, and 5. Although equivalent fluid 
pressures include seepage effects and time-conditioned 
changes in the backfill material, adequate drainage 
should be provided. 



14-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



CROUNO SURFACE 



BACKFILL ffi.C, t) 

VARIATION IN 
ACTIVE EARTH 
PRESSURE 




SURFACE 

<ACKFrLLt«.C,r) 
BACK 




HEEL 



A. GRAVITY 



B. SEMIGRAVITY 



tROUNO SURFACE 



BACKFILL <0,C,f> 




-8ASE 

■"•SO/i PRESSURE 

C. CAT-IT I LEVER 



SURFACE 

£ — BACKFILL <0, C,lt 
COUNTERFORT 




D. COUNTERFORT 



U. S. Army Corps of Engineers 



Figure 14-1. Load diagrams for retaining walls. 



14-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



H 




2 *A 



Where K„ = 



2 

sin (a + <fr) 



sin a sin(a - &) 



E^s? 



6) smU - e 



6) Sin(a + 6 



•'• Component of P. perpendicular to wall back is: 

vH 2 
P AN = P A cos 6 = V K A cos 6 



Special cases 

® If a 



90° , 3=0°, then: 



K A 2 K A 



where K. = 



-.2 



cos <t> 



\cos6 +ysin(6 + 4>) sin <$> 



(Typical K. values for this case are given in Fig. 14-4.) 
(J) If, in addition, 6=0: 



K A = 



cos 



j, = l_ JL _sJr LJU tan 2 {45 _ i } 



(1 + sin 41) 1 + sin <f> 



U. S. Army Corps of Engineers 



Figure 14-2. Active pressure of sand with planar boundaries. 

14-3 



TM 5-818-1 / AFM 88-3, Chap. 7 




p 2 p 



Where K_ = 



sin (a - ♦ ) 



sin a sin(a + 6) 



E - W. 1 1 H : J 



.'.Component of P perpendicular to wall back is: 



P nn = P„ cos 6 = *{}- K cos 5 
pn p 2 p 



Special cases 

© If a = 90°, 6=0°, then: 



p 2 p 



where K_ = 



cos <t> T 

Vcos 5 - VsinU + 6) sin 5 J 



p |_^os 5 

(2) If, in addition, 6=0: 

cosi* = 1 + sin * = Un 2/ 45 + *\ 
- sin *r 1 - sin d> \ */ 



K n = — 

P (1 



(Typical values for this case are given in Fj'g. 14-4.) 

Note: Equations are unconservative and should not be used for 6 > £ ; 
they are satisfactory for 5 s | . J 

U. S. Army Corps of Engineers 

Figure 14-3. Passive pressure of sand with planar boundaries. 



14-4 



TM 5-818-1 / AFM 88-3, Chap. 7 



14-4. Design procedures for retaining walls. 

a. Criteria forselecting earth pressures. 

(1) The equivalent fluid method should be 
used for estimating active earth pressures on retaining 
structures up to 20 feet high, with the addi tion to ea rth 
pressures resulting from backfill compaction Kfiq 14-81 . 

(2) For walls higher than 20 feet, charts, 
equations, or graphical solutions should be used for 
computing lateral earth pressures, with the addition of 
earth pressures resulting from backfill compaction. 

(3) Use at-rest pressures for rigid retaining 
structures resting on rock or batter piles. Design 
cantilever walls founded on rock or restrained from 
lateral movement for at-rest pressures near the base of 
the wall, active pressures along the upper portions of the 
wall, and compaction-induced earth pressures from the 
top to the depth at which they no longer increase lateral 



earth pressures Kfiq 1 4-81 Generally, a linear variation 
in earth pressure coefficients with depth may be 
assumed between the sections of wall. 

(4) Consider passive pressures in the 
design if applied loads force the structure to move 
against the soil. Passive pressures in front of retaining 
walls are partially effective in resisting horizontal sliding. 

b. Overturning. Calculate the factor of safety, 
FS, against overturning, defined as the ratio of resisting 
moments to the overturning moments. Calculate the 
resultant force using load diagrams shown i n] figure 14"T1 
as well as other loadings that may be applicable. Use 
only half of the ultimate passive resistance in calculating 
the safety factor. The resultant of all forces acting on the 
retaining wall should fall within the middle third to provide 
a safety factor with respect to overturning equal to or 
greater than 1.5. 



= 25* 




+8 



10 20 30 

ANGLE OF WALL FRICTION 

A. ACTIVE EARTH PRESSURE 



30« 



O ¥ 
H 20 



W 



u 



I "■ 10° 

J 
J 
< 

* 



1 



I 


•b ,4 


•A 




, ,rf 


r 


i /I 


* / 

V/\ 




<£' 


r 


' •* *.»<■ 


i/i 


// 






SAND^" 




\k 








(C) 





10 15 20 

VALUES OF K„ 



25 



B. PASSIVE EARTH PRESSURE 

(Courtesy of R. B. Peck, W. E. Hanson, and T. H. 
Thornburn, Foundation Engineering , 1 974, p 309. Reprinted 
by permission of John Wiley & Sons. Inc., New York.)\ 4-5 

Figure 14-4. Active and passive earth pressure coefficients according to Coulomb theory. 



14-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



c. Sliding. 

(1) The factor of safety against sliding, 
calculated as the ratio of forces resisting movement to 
the horizontal component of earth plus water pressure on 
the back wall, should be not less than 2.0. If soil in front 
of the toe is disturbed or loses its strength because of 
possible excavation, ponding, or freezing and thawing, 
passive resistance at the toe, P p , should be neglected 
and the minimum factor of safety lowered to 1.5; but if 



the potential maximum passive resistance is small, the 
safety factor should remain at 2.0 or higher. 

(2) For high walls, determine the shearing 
resistance between the base of wall and soil from 
laboratory direct shear tests in which the adhesion 
between the concrete and the undisturbed soil is 
measured. In the absence of tests, the coefficient of 
friction between 



•.a - 



• .4 - 



3 1.4 |- 

< 
> 



1.4 







i 


1 1 


■* 


■>». 




«^& """ 




"•v 


••». 






\ 


•• 


*"^ 




\ 




••- 1 J 




N. 




' / 




LINI \ 




: 1S~' 


_ 


LOAO \ 




jjr _ 


- 


L 




- 


— 


/f 


„ 


■i. 






0.1 


0.40M 




A " / — 


0.3 


0.40H 




/ / / 


0.3 


0.J4M 




// / 


0.7 


0.4IH 1 










iv /ill 


1 


1 1 




VAiuf or 



'« (a,) 



VAtut or » 



■ (£) 



UNI 10 A & O. 



POINT IOAO Q- 




t _ ,n. o.20« 

^^~^ n <«S.' 11 11. . 



Q t ' (0.14 ♦ -')' 



PO« at > 0.4: 



Ms:) 



1.31.'. 



MJUITANT P H . 




<•* » I) 

piisiutcs riOM iini ioao o, 

(IOUSSINISQ IQUATION MOOIPIID IV tXP(IIMCNT) 



U. S. Army Corps of Engineers 



fOt ■ j 0.4 : 

M *Q^' (O.I4»«J) 3 

PO« m > 0.4: 



1.77 










r' H - » H ...*(!. I ») 



sicriON ... 

PltSSUtlS MOM POINT LOAD Q, 
(lOUJSINfiQ (OUATION MOOirifO 
IY IXMIIMINI) 



Figure 14-5. Horizontal pressures on walls due to surcharge. 



14-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



concrete and soil may be taken as 0.55 for 
coarsegrained soils without silt, 0.45 for coarserained 
soils with silt, and 0.35 for silt. The soil in a layer 
beneath the base may be weaker, and the shearing 
resistance between the base of wall and soil should 
never be assumed to exceed the soil strength. Consider 
maximum uplift pressures that may develop beneath the 
base. 



(3) If the factor of safety against sliding is 
insufficient, increase resistance by either increasing the 
width of the base or lowering the base elevation. If the 
wall is founded on clay, the resistance against sliding 
should be based on s u for short-term analysis and ()>' for 
long-term analysis. 



-****$* 




X _ 

k. 3 

O 

M kl 

S» a- 



k. 

. so 



20 



30 



40 



40 



20 



St; 

at 











J 


X 












®, 


/ 


<M 


S.6) 










.+<*&. 




















\K H* 






I 



.- KO 

















k. 

at mo 

3 










^ 


>c 
















Z 






w 

100 








X 

< 
X 










k. 

•» 

■L 

•0 
















X 

• 








IO 




^ 


'>? 


O 













<; 


b 




m 40 








M 












'© 






3 

« 20 










5 *° 
o 




64 

1 




11 
1 


L 

i • 


'*;• 




< 


» 


1 





21 


> 









v ,u ZO *o «n 

VALUES OF SLOPE ANCLE #, D«REES 



80 
60 

40 

20 



160 

MO 

120 

100 

60 
60 
40 
20 
O 



ORCLEO NUMBERS INDICATE THE FOLLOWING SOILS TYPES: 
1. CLEAN SANO AND GRAVEL: GW, GP, SW, SP. 
t DIRTY SAND AND GRAVEL OF RESTRICTED PERMEABILITY: 

CM, GM-GP, SM. SM-SP. 
3. STIFF RESIDUAL SILTS AND CLAYS, SILTY FINE SANOS 

CLAYEY SANOS AND GRAVELS: CL, ML, CH, MH. CM, SC, GC 
i VERY SOFT TO SOFT CLAY, SILTY CLAY. ORGANIC SILT 

AND CLAY: CL, ML, OL, CH. MH. OH. 
f. MEDIUM TO STIFF CLAY DEPOSITED IN CHUNKS AND 

PROTECTED FROM INFILTRATION: CL. CH. 

FOR TYPE 5 MATERIAL H IS REDUCED BY 4 FT, RESULTANT ACTS 
AT A HEIGHT OF (H-4)/3 ABOVE BASE. 



(NA VFAC DM-7) 



Figure 14-6. Design loads for low retaining walls, straight slope backfill. 



14-7 



TM 5-818-1 / AFM 88-3, Chap. 7 



d. Bearing capaci ty. Calcu late from the 
bearing capacity analysis in bhapter 671 Consider local 
building codes or experience where applicable. 

e. Settlement and tilting. When a high 
retaining wall is to be founded on compressible soils, 
estimate total and differential settlements using 



procedures outlined in ( thapter 5~l Reduce excessive 
total settlement by enlarging the base width of the wall or 
by using lightweight backfill material. Reduce tilting 
induced by differential settlement by proportioning the 
size of the base such that the resultant force falls close 



SOIL TYPE I 



son. type z 



SOU. TYPE 3 



x to- 



' K :j2^vfe 










^.-■5 — ^ l 



at 0.4 o* 



ItOi 



OS 100 Q2 0.4 0* 
VALUES OF RATIO Hi/H 



SOIL TYPE 4 



LO O 0.Z 04 Ot O* 10 



TYPE 9 



HO 
120 

x m 

3 

B to 

g to 



40 

to 



MAX SLOPE 3i I 



Ky'O 




& 

.<££- 



:*?> 



£U 



y 



v\. 



••.I 



o 0.x a* o* o.t u> o ot o.< 

VALUES OF RATO H,/M 




at at ix> 



H,«0 

► IJ I/H..K 2 .. 



• i - 



fy 7 ?***!. 



vr 



FOR TYPE S MATERAL H IS REOUCED BY 4 FT, RESULTANT ACTS 

AT A HEIGHT OF W-4J/3 ABOVE BASE. 
FOR DESCRIPTION OF SOIL TYPE SEE FIGURE lk-6 



(NAVFAC DM-7) 



Figure 14-7. Design loads for low retaining walls, broken slope backfill. 



14-8 



TM 5-818-1 / AFM 88-3, Chap. 7 



to the center of the base. Limit differential settlement to 
the amount of tilting that should not exceed 0.05H. If 
settlements are excessive, stabilize compressible soils 
by surcharge loading or a support wall on piles. 

f. Deep-seated failure. Check the overall 
stability of the retaining wall against a deep-seated 
foundation failure using methods of analysis outlined in 
I chapter 81 Forces considered include weight of retaining 
wall, weight of soil, unbalanced water pressure, 
equipment, and future construction. The minimum safety 
factor is 1 .5. 



g. Use of piles. When stability against bearing 
capacity failure cannot be satisfied or settlement is 
excessive, consider a pile foundation. Use batter piles if 
the horizontal thrust of the lateral earth pressure is high. 

14-5. Crib wall. Design criteria of crib walls are 
presented ir l figure 14^91 

14-6. Excavation support systems. The use of 

steep or vertical slopes for a deep excavation is often 
necessitated by land area availability or economics. 




LATERAL PRESSURE IN EXCESS OF 
ACTIVE PRESSURE INDUCED BY 
COMPACTION. 



COMPACTION EQUIPMENT 

10-TON SMOOTH WHEEL ROLLER 
3.2-TON VIBRATORY ROLLER 
1.4-TON VIBRATORY ROLLER 
400-KG VIBRATORY PLATE 
120-KG VIBRATORY PLATE 



K,o v K a v 

CRITICAL 
DEPTH. D c , ft 


<C H > e 


1.9 
1.7 
1.2 
1.5 
1.0 


420 
400 
200 
340 
240 



U. S. Army Corps of Engineers 

Figure 14-8. Estimates of increased pressure induced by compaction. 

14-9 



TM 5-818-1 / AFM 88-3, Chap. 7 



Such slopes are commonly supported by a cantilever 
wall (only for shallow exc avations), a braced wall, or a 
tieback wall ( fig 14-10T71 In some cases, it may be 
economical to mix systems, such as a free slope and a 
tiebac k wall or a tieback wall and a braced wall. ITablel 
114-11 summarizes the wall types with their typical 



prope rties and advantages and disadvantages. I Tablel 
1 14-21 lists factors for selecting wall support system sfor a 
deep excavation (>20 feet). I Table 1 4^31 gives design 
parameters, such as factors of safety, heave problems, 
and supplemental references. 




TYPICAL SECTION 




FISH TAIL TY PE ASSEMBLY 



^filler ktmck 

'Anchor titk 




CLOSE FACE ASSEMBLY 




CORNER. OF BIN ASSEMBLY 



CRIB RETAINING WALLS 

TYPES • Commofl type* of cribs show* 

«a *ecemp*B.yl&f diagram*. 



CRIBBING MATERIALS 

crcte, and metal. 



Timber, con* 



FILL - Crushed stone, ether coaree 
granular material. Including rock lea* 
than U-ln. *l*«. 

DESIGN - Design criteria for gravity 
walle »pply. Wall section restating 

overturning is taken as a rectangle of 
dimension (II x b). 

Weight of crib is equal to that of mate- 
rial within (H x b), including weight of 
cr£b members. 

Low walls (4 ft High and under) may be 
mad* with a plumb face. Higher walls 
are battered on the face at least I in. 
per foot. For biiK walls (l£ it High 
and over] the batter ie increased or *up- 
plemeatal cribs added at tbc back. 
IflTopco face cribs, the space between 
stretchers should not exceed ft in. so a* 
to properly retain the fill. 
Expansion joints for concrete aad metal 
cribbing are spaced no more than °0 ft* 

FILLING • The wall should not be laid 
up higher than 3 ft above the level of the 
fill within the crib. 

DRAINAGE * FROST ACTION - No 
special requirements, wall should bo 
made free draining. 



BIN TYJ>£ RETAINING WALL - Corn- 
posed of metal bin* or cell* joined to 
special columnar unit* at the comer*. 
The design requirements are the same 
aa for crib walla except that auitable 
drainage behind the wall is needed. 



(NAVFAC DM-7) 



Figure 14-9. Design criteria for crib and bin walls. 



14-10 



TM 5-818-1 / AFM 88-3, Chap. 7 



14-7. Strutted excavations. 

a. Empirical design earth press ure diagrams 
developed from observations are shown in l figure 14-11 J 
In soft to medium clays, a value of m = 1.0 should be 
applied if a stiff stratum is present at or near the base of 
the excavation. If the soft material extends to a sufficient 
depth below the bottom of the excavation and significant 
plastic yielding occurs, a value of 0.4 should be used for 



wmwM. 



m. The degree of plastic yielding beneath an excavation 
is governed by the stability number N expressed as 



N = yH/s u 



(14-2) 



where y, H, and s u , are defined in Ifigure 14-11.1 If N 
exceeds about 4, m < 1 .0. 



'mmm 



; ^zzzzzzzzzzzzfr / 



r 



STRUTS 



yiruritnn^ J 



' tzzzzzzzzzzzzd ' 



, mmm?. w^y h.'^ 



CANTILEVER WALL 



b. CROSS-LOT BRACED WALL 



RAKERS 




■KICKER BLOCK OR 
FOUNDATION SLAB 



ANCHORS 




RAKER SYSTEM 



ANCHOR OR TIEBACK WALL 





e. FREE SLOPE-TIEBACK WALL 
U. S. Army Corps of Engineers 



f. TIEBACK-BRACED WALL 



Figure 14-10. Types of support systems for excavation. 



14-11 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 14-1. Types of Walls 



Name 

Steel sheeting 



Section 



Typical El 

Values per foot 

ksf 

900-90, 000 



Soldier pile 
and lagging 



Cast-in-place 
concrete 
slurry wall 



2,000-120,000 



288, 000- 
2, 300, 000 



Precast con- 
crete slurry 



288,000- 
2,300,000 





Advantages 


(1) 


Disadvantages 


(1) 


Can be impervious 


Limited stiffness 


(2) 


Easy to handle and 


(2) 


Interlocks can be 




construct 




lost in hard driv- 


(3) 


Low initial cost 




ing or in gravelly 
soils 


(1) 


Easy to handle and 


(1) 


Wall is pervious 




construct 


(2) 


Requires care in 


(2) 


Low initial cost 




placement of 


(3) 


Can be driven or 
augered 




lagging 


(1) 


Can be impervious 


(1) 


High initial cost 


(2) 


High stiffness 


(2) 


Specialty contractor 


(3) 


Can be part of 




required to 




permanent 




construct 




structure 


(3) 


Extensive slurry 
disposal needed 






(4) 


Surface can be 
rough. 


(1) 


Can be impervious 


(1) 


High initial cost 


(2) 


High stiffness 


(2) 


Specialty contractor 


(3) 


Can be part of 




required to 




permanent 




construct 




structure 


(3) 


Slurry disposal 


(4) 


Can be prestressed 




needed 






(4) 


Very large and 
heavy members must 
be handled for deep 
systems 






(5) 


Permits some 
yielding of subsoils 



(Continued) 



14-12 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 14-1. Types of Walls-Continued 



Typical El 

Name 

Cylinder pile 
wall 



Section 



Values per 


foot 








ksf 




Advantages 


(1) 


Disadvantages 


115,000- 




(1) Secant piles 


High initial cost 


1,000,000 




impervious 

(2) High stiffness 

(3) Highly specialized 
equipment not 
needed for tangent 
piles 

(4) Slurry not needed 


(2) 


Secant piles require 
special equipment 



14-13 



TM 5-818-1 / AFM 88-3, Chap. 7 



b. For stiff-fissured clays, diagram (c) o f fiaurel 
1 14-111 applies for any value of N. If soft clays, diagram 

(b) applies except when the computed maximum 
pressure falls below the value of the maximum pressure 
in diagram (c). In these cases, generally for N < 5 or 6, 
diagram (c) is used as a lower limit. There are no design 
rules for stiff intact clays and for soils characterized by 
both c and (j) such as sandy clays, clayey sands, or 
cohesive silts. 

c. The upper tier of bracing should always be 
installed near the top of the cut, although computations 
may indicate that it could be installed at a greater depth. 
Its location should not exceed 2s u below the top of the 
wall. 

d. Unbalanced water pressures should be 
added to the earth pressures where the water can move 
freely through the soil during the life of the excavation. 
Buoyant unit weight is used for the soil below water. 
Where undrained behavior of a soil is considered to ap- 



ply, the use of total unit weights in calculating earth 
pressures automat ically accoun ts for the loads produced 
by groundwater ( flg 14-1 1"T1 Pressures due to the 
surcharge load are computed as indicated in previous 
sections and added to the earth and water pressures. 

e. Each strut is assumed to support an area 
extending halfway to the adjacent strut ( fig 14-11)1 The 
strut load is obtained by summing the pressure over the 
corresponding tributary area. Temperature effects, such 
as temperature increase or freezing of the retained 
material, may significantly increase strut loads. 

f. Support is carried to the sheeting between 
the struts by horizontal structural members (wales). The 
wale members should be designed to support a 
uniformly distributed lo ad equal to the maximum 
pressure determined fron j figure 14-11 I times the spacing 
between the wales. The wales may be assumed to be 
simply supported (pinned) at the struts. 



Table 14-2. Factors Involved in Choice of a Support System for a Deep Excavation 







Lends Itself To Use 


Downgrades Utility 






Requirement 


Of 


Of 


Comment 


1. 


Open excavation area 


Tiebacks or rakers or 
cantilever walls (shal- 
low excavation) 


Crosslot struts 




2. 


Low initial cost 


Soldier pile or sheetpile 
walls; combined soil 
slope with wall 


Diaphragm walls, cyl- 
inder pile walls 


Depends somewhat on 3 


3. 


Use as part of per- 


Diaphragm or cylinder 


Sheetpile or soldier 


Diaphragm wall most com- 




manent structure 


pile walls 


pile walls 


mon as permanent 
wall 
Tieback capacity not 


4. 


Deep, soft clay sub- 


Strutted or raker sup- 


Tiebacks, flexible 




surface conditions 


ported diaphragm or 
cylinder pile walls 


walls 


adequate in soft clays 


5. 


Dense, gravelly sand 


Soldier pile, diaphragm 


Sheetpile walls 


Sheetpile walls lose in- 






or clay subsoils 


or cylinder pile 


terlock on hard driving 


6. 


Deep, overconsoli- 


Struts, long tiebacks 


Short tiebacks 


High lateral stresses are 




dated clays 


or combination tie- 
backs and struts 
l/tia. 14-101 




relieved in O.C. soils 
and lateral movements 
may be large and 
extend deep into soil 


7. 


Avoid dewatering 


Diaphragm walls, pos- 
sibly sheetpile walls 
in soft subsoils 


Soldier pile wall 


Soldier pile wall is 
pervious 


8. 


Minimize movements 


High preloads on stiff 
strutted or tied-back 
wall 

Tiebacks or rakers 


Flexible walls 


Analyze for stability of 
bottom of excavation 


9. 


Wide excavation 


Crosslot struts 


Tiebacks preferable ex- 




(greater than 20m 






cept in very soft clay 




wide) 






subsoils 


10. 


Narrow excava- 
tion (less than 
20m wide) 


Crosslot struts 


Tiebacks or rakers 


Struts more economical 
but tiebacks still may 
be preferred to keep 
excavation open 



U. S. Army Corps of Engineers 



14-14 



Design Factor 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 14-3. Design Considerations for Braced and Tieback Walls 

Comments 



1 . Earth loads 



2. Water loads 



3. Stability 



4. Piping 



5. Movements 



6. Dewatering - recharge 



7. Surcharge 



For struts, select from the semiempirical diaaram sRfia. 14-10 ): for walls 
and wales use lower loads - reduce by 25 percent from strut loading. 
Tiebacks may be designed for lower loads than struts unless preloaded to 
higher values to reduce movements 

Often greater than earth load on impervious wall. Should consider possi- 
ble lower water pressures as a result of seepage through or under wall. 
Dewatering can be used to reduce water loads 

Consider possible instability in any berm or exposed slope.Sliding po- 
tential beneath the wall or behind tiebacks should be evaluated. Deep 
seated bearing failure under weight of supported soil to be checked in 
soft cohesive soils [fig. 14-121 

Loss of ground caused by high groundwater tables and silty soils. 

Difficulties occur due to flow beneath wall, through bad joints in wall, 
or through unsealed sheetpile handling holes. Dewatering may be 
required. 

Movements can be minimized through use of stiff impervious wall supported 
by preloaded tieb ack or brac ed system. Preloads should be at the level 
of load diagrams Iffig. 14-11 ) for minimizing movements 

Dewatering reduces loads on wall systems and minimizes possible loss of 
ground due to piping. May cause settlements and will then need to re- 
charge outside of support system. Not applicable in clayey soils 

Storage of construction materials usually carried out near wall systems. 
Allowance should always be made for surcharge, especially in upper 
members 

(Continued) 



14-15 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 14-3. Design Considerations for Braced and Tieback Walls-Continued 



Design Factor 



Comments 



8. Preloading 

9. Construction sequence 

10. Temperature 

1 1 . Frost penetration 

12. Earthquakes 

13. Codes 

1 4. Factors of safety 



Useful to remove slack from system and minimize soil movements. Preload 
up to the load diagram loads [fig. 14-101 to minimize movements 

Sequence used to build wall important in loads and movements of system. 
Moments in walls should be checked at every major construction stage 
for maximum condition. Upper struts should be installed early 

Struts subject to load fluctuation due to temperature loads; may be impor- 
tant for long struts 

In very cold climates, frost penetration can cause significant loading on 
wall system. Design of upper portion of system should be conservative. 
Anchors may have to be heated 

Seismic loads may be induced during earthquake. Local codes often govern 

For shallow excavations, codes completely specify support system. Varies 
from locality to locality. Consult OSHA requirements 

Minimum Design 
Item Factor of Safety 



Earth Berms 
Critical Slopes 
Noncritical Slopes 
Basal Heave 
General Stability 



2.0 
1.5 
1.2 
1.5 

1.5 



14-16 



TM 5-818-1 / AFM 88-3, Chap. 7 



14-8. Stability of bottom of excavation. 

a. Piping in sand. The base of an excavation 
in sand is usually stable unless an unbalanced 
hydrostatic head creates a "quick" condition. Among the 
methods to eliminate instability are dewatering, 
application of a surcharge load at the bottom of the 
excavation, and deeper penetration of the piling. 

b. Heaving in clays. The stability against 
heave of the b ottom of an e xcavation in soft clay may be 
evaluated fror rl figure 14-13 - If the factor of safety is less 
than 1 .5, the piling should be extended below the base of 
the excavation. Heave may also occur because of 



unrelieved hydrostatic pressures in a permeable layer 
located below the clay. 

c. Care of seepage. Small amounts of 
seepage into the excavation can be controlled by 
pumping from sumps. Such seepage can be expected if 
the excavation extends below the water table into 
permeable soils. If the soils consist of fine sands and 
silts, the sumps should be routinely monitored for 
evidence of fines being washed from the soil by 
seepage. If large quantities of fine-grained materials are 
found in the sumps, precautionary steps should be taken 
to make the lagging or sheeting watertight to avoid 
excessive settlements adjacent to the excavation. 



(o) SANDS 



(b) SOFT TO FIRM CLAYS 



(e) STIFF FISSURED CLAYS 



77777777 





T1HWT 



*^\. 0.25H 


* ^^ "*■ 


* 




« 


1 

0.5H H 


^ 


\ 


5^^ 


0.25H 



yH-4m.s„ 



0.2rH TO 0.4-yH 



NOTES : 

1. CHECK SYSTEM FOR PARTIAL EXCAVATION CONDITION 

2. IF THE FREE WATER LEVEL IS ABOVE THE BASE OF THE EXCAVATION THE 
HYDROSTATIC PRESSURE MUST BE ADDED TO THE ABOVE PRESSURE 
DISTRIBUTION IN SANDS 

3. IF SURCHARGE LOADINGS ARE PRESENT AT OR NEAR THE GROUND 
SURFACE THESE MUST BE INCLUDED IN THE LATERAL PRESSURE 
CALCULATION . 

4. VALUES OF m ARE GIVEN IN PARAGRAPH 14-7. 

5. ' y= UNIT WEIGHT OF SOIL. 

S u = UNDRAINED SHEAR STRENGTH. 



U. S. Army Corps of Engineers 



Figure 14-11. Pressure distribution-complete excavation. 



14-17 



TM 8-818-1 / AFM 88-3, Chap. 7 




q=7H+ q 

1W 



CN, 



D>0.7B F.S. 



N FROM I 




SQUARE OR ~ 
CIRCULAR 
BASE 
B/L -0.0 

INFINITELY LONG 
BASE B/L =0.0 



J I I L 



1 2 3 4 5 6 

H/B 

:n c = (, + o. 2 f)N c jn(# 



RECTANGULAR BASE I 

WHERE L = LENGTH OF EXCAVATION 




CN, 



D <0.7B F.S. 



U. S. Army Corps of Engineers 



<n 1.3 




LONG EXCAVATION: 

N cinf. = f ( D/BH/3 ) = K ) K 2 

RECTANGULAR EXCAVATION: 

N c = ( 1+02 l) N cinf. 



Figure 14-12. Stability of bottom of excavation in clay. 



14-18 



TM 5-818-1 / AFM 88-3, Chap. 7 



14-9. Anchored walls. 

a. Tiebacks have supplanted both strut and 
raker systems in man y instance s to support wide 
excavations. The tieback lTrTa 14-13 ) connects the wall to 
an anchorage located in a zone where significant soil 
movements do not occur. The anchorage may be in soil 
or rock; soft clays probably present the only condition 
w here an ancho rage in soil cannot be obtained reliably. 
In Ifiqure 14-13, I the distance L ub should extend beyond 
the "Rankine" zone some distance. This distance is 
necessary, in part, to obtain sufficient elongation in 
anchored length of rod L a during jacking so that soil 
creep leaves sufficient elongation that the design load is 
retained in the tendon. After jacking, if the soil is 
corrosive and the excavation is open for a long time, the 
zone L ub may be grouted. Alternatively, the length of 
tendon L ub is painted or wrapped with a grease 
impregnated wrapper (prior to placing in position). 

b. The tieback tendon may be either a single 
high-strength bar or several high-strength cables (fy on 



the order of 200 to 270 kips per square inch) bunched. It 
is usually inclined so as to reach better bearing material, 
to avoid hole collapse during drilling, and to pass under 
utilities. Since only the horizontal component of the 
tendon force holds the wall, the tendon should be 
inclined a minimum. 

c. Tieback anchorages may be drilled using 
continuous flight earth augers (commonly 4 to 7 inches in 
diameter) and may require casin g to hold the hole until 
grout is placed in the zone L a of l figure 14-131 at which 
time the casing is withdrawn. Grout is commonly used 
under a pressure ranging from 5 to 150 pounds per 
square inch. Underreaming may be used to increase the 
anchor capacity in cohesive soil. Belling is not possible 
in cohesionless soils because of hole caving. Typical 
formulas that can be used to compute the capacity of 
tieback anchorages are given ir | figure 14-141 

d. Exact knowledge of the anchor capacity is 
not 




PROTECTIVE JACKET 
(AS REQUIRED) 



U. S. Army Corps of Engineers 



Figure 14-13. Typical tieback details. 
14-19 



TM 5-818-1 / AFM 88-3, Chap. 7 



Case 1 ■ 


- Straight Shaft Anchors 




K 


^Sw >^ unbonded | 
^SS. d 










mmmm 



Q = TrB(c, + o„ tan 6) L a ; c a = adhesion on shaft 
u an a a 



c a = s,, in clay with s„ < 0.5 tsf 
a u u 

c_ = 0.5 tsf in clay with s„ > 0.5 tsf 
a u 

c, = in sand 

a 

6 = angle of frictional 
resistance at grout- 
soil interface 
commonly 6 ^ <j, 

B = diameter of hole for tremie or if hole was cased 

= diameter computed from grout volume and L as 

a 



"•^0.7854 L. 



a = normal stress on center of anchor; 
determine as per Mohr circle. If 
grouted anchor, a may be higher 

Vg = volume of grout in length L 




Case 2 - Underreamed Anchor (Clays Only) 

Drill case embedded 

to form seal-^ , i 

\ . h L rr— 

Adhesion on shaft ' 




|B(diam. of underream) 
Shear in clay along underream 



Q U = irBL 2 S u F l +J<B 2 -d 2 )N c S u + 1 rdLc a 



F, ranges from 1 to 0.75 depending upon amount of disturbance 
N = 9 (range of 5.7 to 9 depending on depth) 



U. S. Army Corps of Engineers 



Figure 14-14. Methods of calculating anchor capacities in soil. 



14-20 



TM 5-818-1 / AFM 88-3, Chap. 7 



needed as all the anchors are effectively "proof-tested" 
(about 120 to 150 percent of design load) when the 
tendons are tensioned for the design load. One or more 
anchors may be loaded to failure; however, as the cost 
of replacing a failed anchor is often two to three times 
the cost of an initial insertion, care should be taken not 



to fail a large number of anchors in any-test program. If 
the tieback extends into the property of others, 
permission, and possibly a fee, will be required. The 
tieback tendons and anchorages should normally be left 
in situ after construction is completed. Se e table 14-3 for 
additional design considerations. 



14-21 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 15 
FOUNDATIONS ON FILL AND BACKFILLING 



15-1. Types of fill. 

a. Fills include conventional compacted fills; 
hydraulic fills; and uncontrolled fills of soils or industrial 
and domestic wastes, such as ashes, slag, chemical 
wastes, building rubble, and refuse. Properly placed 
compacted fill will be more rigid and uniform and have 
greater strength than most natural soils. Hydraulic fills 
may be compacted or uncompacted and are an 
economical means of providing fill over large areas. 
Except when cohesionless materials, i.e., clean sands 
and gravels, are placed under controlled conditions so 
silty pockets are avoided and are compacted as they are 
placed, hydraulic fills will generally require some type of 
stabilization to ensure adequate foundations. 

b. Uncontrolled fills are likely to provide a 
variable bearing capacity and result in a nonuniform 
settlement. They may contain injurious chemicals and, 
in some instances, may be chemically active and 
generate gases that must be conducted away from the 
structure. Foundations on fills of the second and third 
groups (and the first group if not adequately compacted) 
should be subjected to detailed investigations to 
determine their suitability for supporting a structure, or 
else they should be avoided. Unsuitable fills often can 
be adequately stabilized. 

15-2. Foundations on compacted fills. 

a. Compacted fill beneath foundations. 
Compacted fills are used beneath foundations where it is 
necessary to raise the grade of the structure above 
existing ground or to replace unsatisfactory surface soils. 
Fills constructed above the natural ground surface 
increase the load on underlying soils, causing larger 
settlements unless construction of the structure is 
postponed until fill-induced settlements have taken 
place. Settlements beneath a propo sed fill can be 
computed using methods outlined in Ichapter 5~l If 
computed settlements are excessive, consider 
surcharging and postponing construction until the 
expected settlement under the permanent fill loading has 
occurred. Extend the fill well beyond the loading area, 
except where the fill is placed against a cut slope. 
Where the fill is relatively thick and is underlain by soft 
materials, check its stability with respect to deep sliding. 
If the fill is underlain by weaker materials, found the 
footings on the fill unless settlement is excessive. If the 
fill is underlain by a stronger material, the footings may 
be founded on the fill or on the stronger material. 



b. Foundations partially on fill. Where a 
Sloping ground surface or variable foundation depths 
would result in supporting a foundation partially on 
natural soil, or rock, and partially on compacted fill, 
settlement analyses are required to estimate differential 
settlements. In general, a vertical joint in the structure 
should be provided, with suitable architectural treatment, 
at the juncture between the different segments of 
foundations. The subgrade beneath the portions of 
foundations to be supported on natural soils or rock 
should be undercut about 3 feet and replaced by 
compacted fill that is placed at the same time as the fill 
for the portions to be supported on thicker compacted fill. 

c. Location of borrow. Exploratory 
investigations should be made to determine the suitable 
sources of borrow material. Laboratory tests to 
determine the suitability of available materials include 
natural water contents, compaction characteristics, grain- 
size distribution, Atterberg limits, shear strength, and 
consolidation. Typical properties of compacte d materials 
for use in preliminary analyses are given in Itable 3^] 
The susceptibility to frost action also should be 
considered in analyzing the potential behavior of fill 
material. The scope of laboratory testing on compacted 
samples depends on the size and cost of the structure, 
thickness and extent of the fill, and also strength and 
compressibility of underlying soils. Coarse-grained soils 
are preferred for fill; however, most fine-grained soils can 
be used advantageously if attention is given to drainage, 
compaction requirements, compaction moisture, and 
density control. 

d. Design of foundations on fill. Foundations 
can be designed on the basis of be aring capaci ty and 
settlement calculations described in chapter 10l The 
settlement and bearing capacity of underlying foundation 
soils also should be evaluated. Practically all types of 
construction can be founded on compacted fills, 
provided the structure is designed to tolerate anticipated 
settlements and the fill is properly placed and 
compacted. Good and continuous field inspection is 
essential. 

e. Site preparation. The site should be 
prepared by clearing and grubbing all grass, trees, 
shrubs, etc. Save as many trees as possible for 
environmental considerations. The topsoil should be 
stripped and stockpiled for later landscaping of fill and 
borrow areas. Placing and compacting fills should 
preferably be done when 



15-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



the area is still unobstructed by footings or other 
construction. The adequacy of compacted fills for 
supporting structures is dependent chiefly on the 
uniformity of the compaction effort. Compaction 
equipment generally can be used economically and 
efficiently only on large areas. Adverse weather 
conditions may have a pronounced effect on the cost of 
compacted fills that are sensitive to placement moisture 
content, i.e., on materials having more than 10 to 20 
percent finer than the No. 200 sieve, depending on 
gradation. 

f. Site problems. Small building areas or 
congested areas where many small buildings or utility 
lines surround the site present difficulties in regard to 
maneuvering large compaction equipment. Backfilling 
adjacent to structures also presents difficulties, and 
power hand-tamping equipment must be employed, with 
considerable care necessary to secure uniform 
compaction. Procedures for backfilling a round struct ures 
are discussed in TM 5-818-4 / AFM 88-5, IChapteT51 

15-3. Compaction requirements. 

a. General. Guidelines for selecting 
compaction equipment and for establishing compaction 
requirements for various soil types are given in table 15- 
1 . If fill materials have been thoroughly investigated and 
there is ample local experience in compacting them, it is 
preferable to specify details of compaction procedures, 
such as placement water content, lift thickness, type of 
equipment, and number of passes. When the source of 
the fill or the type of compaction equipment is not known 
beforehand, specifications should be based on the 
desired compaction result, with a specified minimum 
number of coverages of suitable equipment to assure 
uniformity of compacted densities. 

b. Compaction specifications. For most 
projects the placement water content of soils sensitive to 
compaction moisture should be within the range of -1 to 
+ 2 percent of optimum water content for the field 
compaction effort applied. Each layer is compacted to 
not less th an the perc entage of maximum density 
specified in | table 15-2| It is generally important to 
specify a high degree of compaction in fills under 
structures to minimize settlement and to ensure stability 
of a structure. I n addition t o criteria set forth in TM 5- 
818-4/AFM 88-5. IChapteT5l the following factors should 
be considered in establishing specific requirements: 

(1) The sensitivity of the structure to total 
and differential settlement as related to structural design 
is particularly characteristic of structures to be founded 
partly on fill and partly on natural ground. 

(2) If the ability of normal compaction 
equipment to produce desired densities in existing or 
locally available materials within a reasonable range of 
placement water content is considered essential, special 
equipment should be specified. 



(3) The compaction requirements for 
clean, cohesionless, granular materials will be generally 
higher than those for cohesive materials, because 
cohesionless materials readily consolidate, or liquify, 
when subjected to vibration. For structures with unusual 
stability requirements and settlement limit ations, the 
minimum density requirements indicated in I table 15-21 
should be increased. For coarse-grained, well-graded, 
cohesionless soils with less than 4 percent passing the 
No. 200 sieve, or for poorly graded cohesionless soils 
with less than 10 percent, the material should be 
compacted at the highest practical water content, 
preferably saturated. Compaction by vibratory rollers 
generally is the most effective procedure. Experience 
indicates that pervious materials can be compacted to an 
average relative density of 85 + 5 percent with no 
practical difficulty. For cohesionless materials, stipulate 
that the fill be compacted to either a minimum density of 
85 percent relative density or 95 percent of CE 55 
compaction effort, whichever gives the greater density. 

(4) If it is necessary to use fill material 
having a tendency to swell, the material should be 
compacted at water contents somewhat higher than 
optimum and to no greater density than require d for 
stability under proposed loadings (l iable 15-2). I The 
bearing capacity and settlement characteristics of the fill 
under these conditions should be checked by laboratory 
tests and analysis. Swelling clays can, in some 
instances, be permanently transformed into soils of lower 
plasticity and swelling potential by adding a small 
percentage of hydrated lime (chap 16). 

c. Compacted rock. Compacted crushed rock 
provides an excellent foundation fill. Vibratory rollers are 
preferable for compacting rock. Settlement of fill under 
the action of the roller provides the most useful 
information for determining the proper loose lift 
thickness, number of passes, roller type, and material 
gradation. Compaction with a 10-ton vibratory roller is 
generally preferable. The rock should be kept watered at 
all times during compaction to obviate collapse 
settlement on loading and first wetting. As general 
criteria for construction and control testing of 
rock fill are not available, test fills should be employed 
where previous experience is inadequate and for large 
important rock fills. 

15-4. Placing and control of backfill. Backfill should 
be placed in lifts no greater than shown in I table 15TI 
preferably 8 inches or less and dependingon the soil and 
type of equipment available. No backfill should be 
placed that contains frozen lumps of soil, as later 
thawing will produce local soft spots. Backfill should not 
be placed on muddy, frozen, or frost-cov- 



15-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



Table 15-1. A summary of Density Methods for Building Foundation 



Soil 
Group 


Soil 






fill and Backfill 


Deep Foundation Deposits 


of 
Compaction 


Typical Equipment and Procedures for Compaction 


Field Control 


Compaction 
Methods 


Field Control 


Equ i pment 


No- of Passes 
or Coveragee 


Comp. Lift 
Thick-, in. 


Placement 
Utter Content 


I 

c 

e 

£ 

V 

£ 
3 

O 

* 


OF 
SV 
3? 


1 


90 to 95* of 
"K 5^ maximum 
density 

75 to 85* of 

relative 

density 


Vitratcry mllerii and 
CLT.i.-autcrs 


Indefinite 


Indefinite 


Saturate by flooding 


Control samples at in- 
tervals to determine 
degree of compact ion 
or relative density 


None available 
except for near 
surface (to 
approximate 
depth of 5 rt) 
compaction by 
equipment and 
procedure shown 
at left 




(a) 
Rubber-tired roller 1 ' 


2-5 coverages 


12 


(p) 
Cravler-type tractor 


2-5 coverages 


8 


Power hand tapper' ' 


Indefinite 


6 


■a 

V 

*> 

1 


85 to 90* of 
Ct 5f maximum 
density 

65 to 75* of 

relative 

density 


Rubber-tired roller' 8 ^ 


2-5 coverages 


Ik 


Saturate by flooding 


Control samples as 
noted above, if needed 


Vibroflotatlon, 
compaction piles, 
Band piles, ex- 
plosives 

Surface compac- 
tion as noted 
above 


Undisturbed 

samples from 
boring* or test 
pits to deter- 
mine degree of 
compaction or 
relative density 


Crawler-type tractor* ' 


1-2 coverages 


10 


tc) 
rower hand temper* ' 


Indefinite 


8 


Controlled routing of con- 
•tructlon equipment 


Indefinite 


3-10 


I 

t 

I 

1 

s 
1 

3 


CM 
QC 
91 

sc 

ML 
CL 
01. 
OH 
MB 
CH 
OB 


1 


90 to 95* of 
CE 55 maximum 
density 


Rubber-tired roller*" ' 


2-5 coverages 


8 


Optimum water con- 
tent based on CE 55 
test with 12 blows 
per layer 


Control sample* at in- 
tervals to determine 
degree of compaction 


(A.) Surface compaction by equip- 
ment and procedures shown at left 
Is feasible only if materiel Is at 
proper water content 

(B) Den* If lest Ion of Mil* 1* con- 
trolled by consolidation process 

(a j preload fills* 

(b) lowering of ground-water 
table 

{«.) drying 

* Consolidation amy he ac- 
celerated by Men* of sand 
drains 

field control exercleed by oto- 
eervetlon of pore pressure* and 
surface settlements 


Sheepsfoot roller* 


k-6 passes 


6 


rower hand tamper * c ' 


Indefinite 


k 


5 
1 

I 


85 to 90* of 
CE 55 maximum 
density 


Rubber-tired roller**' 


2-1* coverage* 


10 


(A) Optimum water 
content based on 
CE 55 test with 

7 blows oer layer 

(B) By observation; 
wet e lie -maximum wa- 
ter content at which 
material can eat la* 
factorlly operate, 
dry • Id* -minimum wa- 
ter content required 
to bond particle* 
and which will not 
result in voids or 
DoneycoBbed material 


(A) Control eamplee as 
noted above. If needed 

(B) yield control exer- 
cised by visual Inspec- 
tion of action of com- 
pacting eo^iipment 


(d) 
Sheepsfoot roller* 


U-8 paaaea 


8 


(b) 
Crnvier-type tractor s 


3 coverages 


6 


Power baud tamper^"' 


Indefinite 


6 


Controlled routing of con- 
struction eojiipnent 


Indefinite 


6-8 







The above requirements will be adequate in relation to neat construction. In special cases where tolerable settlements are unusually small, it may be necessary to employ 
additional compaction equivalent to 95 to 100* of CI 55 compaction effort. A coverage consists of one application of the wheel of a rubber-tired roller or the threads of 
s crawler-type tractor over each point In the arts being ccapacted. For a sheepsfoot roller, one put consists of one movement of a sheepsfoot roller drum over the area 
being compacted. 

) Rubber-tired rollers bavins a wheel load between 16,000 and 25,000 lb and a tire pressure between 80 and 10t pel. 

1 Cravler-type tractors weighing not less than 20,000 lb and exerting a foot pressure not leas than 6-1/2 pel. 

J Power hand tampers weighing sore than 1001b; pneumatic or operated by gasoline engine. 

) Sheepsfoot rollers having a foot pressure between 250 and 500 pel and tamping feet 7 to 10 in. in length with a face area between 7 end l6 so. In. 



U. S. Army Corps of Engineers 



15-3 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 15-2. Compaction Density as a Percent of CE 55 Labortory Test Density 



Fill, embankment, and backfill 
Under proposed structures, building 
slabs, steps, and paved areas 

Under sidewalks and grassed areas 

Subqrade 

Under building slabs, steps, and paved 
areas, top 12 in. 

Under sidewalks, top 6 in. 



CE 55 Maximum Density, % 
Cohesive Cohesionless 

Soils Soils 



90 



85 



90 



85 



95 a 



90 



95 



90 



May be 85% relative density, whichever is higher. 

ered ground. Methods of compaction controf during 
constructio n are described in TM 5-818-4/AFM 88-5, 
I Chapter 5J 

15-5. Fill settlements. A fill thickness of even 3 feet 
is a considerable soil load, which will increase stresses 
to a substantial depth (approximately 2B, where B = 
smallest lateral dimension of the fill). Stress increases 
from the fill may be larger than those from structure 
footings p laced on the fill. Use procedures outlined in 
I chapter fb to obtain expected settlements caused by fill 
loading. Many fills are of variable thickness, especially 
where an area is landscaped via both cutting and filling 
to obtain a construction site. In similar cases, attention 
should be given to building locations with respect to 
crossing cut and fill lines so that the proper type of- 
building settlement can be designed (building may act as 
a cantilever, or one end tends to break off, or as a beam 
where the interior sags). Proper placing of reinforcing 
steel in the wall footings (top for cantilever action or 
bottom for simple beam action) may help control building 
cracks where settlement is inevitable; building joints can 
be provided at critical locations if necessary. The 
combined effect of structure (one- and two-story 
residences) and fill loading for fills up to 10 feet in 
thickness on sound soil and using compaction control 
should not produce a differential settlement of either a 
smooth curved hump or sag of 1 inch in 50 feet or a 
uniform slope of 2 inches in 50 feet. 

15-6. Hydraulic fills. Hydraulic fills are placed on 
land or underwater by pumping material through a 
pipeline from a dredge or by bottom dumping from 
barges. Dredge materials vary from sands to silts and 
fine-grained silty clays and clays. Extensive 

maintenance dredging in the United States has resulted 
in disposal areas for dredge materials, which are 



especially attractive from an economic standpoint for 
development purposes. Dikes are usually required to 
retain hydraulic fills on land and may be feasible for 
underwater fills. Underwater dikes may be constructed 
of large stones and gravel. 

a. Pervious fills. Hydraulically placed pervious 
fills with less than 10 percent fines will generally be at a 
relative density of 50 to 60 percent but locally may be 
lower. Controlled placement is necessary to avoid silt 
concentrations. Compaction can be used to pr oduce 
relati ve densities sufficient for foundation support ftablel 

□73}. Existing uncompacted hydraulic fills of pervious 
materials in seismic areas are subject to liquefaction, 
and densification will be required if important structures 
are to be founded on such deposits. Rough estimates of 
relative density may be obtained using standard 
penetration resistance. Undisturbed borings will be 
required to obtain more precise evaluation of in situ 
density and to obtain undisturbed samples for cyclic 
triaxial testing, if required. For new fills, the coarsest 
materials economically available should be used. Unless 
special provisions are made for removal of fines, borrow 
containing more than 10 percent fines passing the No. 
200 sieve should be avoided, and even then controlled 
placement is necessary to avoid local silt concentrations. 

b. Fine-grained. fills. Hydraulically placed 
overconsolidated clays excavated by suction dredges 
produce a fill of clay balls if fines in the washwater are 
permitted to run off. The slope of such fills will be 
extremely flat ranging from about 1 2 to 1 6H on 1 V. 

(1) These fills will undergo large 
immediate con- 



15-4 



TM 5-818-1 / AFM 88-3, Chap. 7 



solidation for about the first 6 months until the clay balls 
distort to close void spaces. Additional settlements for a 
1-year period after this time will total about 3 to 5 percent 
of the fill height. 

(2) Maintenance dredgings and 

hydraulically placed normally consolidated clays will 
initially be at water contents between 4 and 5 times the 
liquid limit. Depending on measures taken to induce 
surface drainage, it will take approximately 2 years 
before a crust is formed sufficient to support light 
equipment and the water content of the underlying 
materials approaches the liquid limit. Placing 1 to 3 feet 
of additional conhesionless borrow can be used to 
improve these areas rapidly so that they can support 
surcharge fills, with or without vertical sand drains to 
accelerate consolidation. After consolidation, substantial 
one- or two-story buildings and spread foundations can 
be used without objectionable settlement. Considerable 
care must be used in applying the surcharge so that the 
shear strength of the soil is not exceeded (i.e., use light 
equipment). 

c. Settlements of hydraulic fills. If the 
coefficient of permeability of a hydraulic fill is less than 
0.0002 foot per minute, the consolidation time for the fill 
will be long and prediction of the behavior of the 
completed fill will be difficult. For coarse-grained 
materials with a larger coefficient of permeability, fill 
consolidation and strength buildup will be relatively rapid 
and reasonable strength estimates can be made. Where 
fill and foundation soils are fine-grained with a low 
coefficient of permeability, piezometers should be placed 
both in the fill and in the underlying soil to monitor 



pore pressure dissipation. It may also be necessary to 
place settlement plates to monitor the settlement. 
Depending on the thickness of the fill, settlement plates 
may be placed both on the underlying soil and within the 
fill to observe settlement rates and amounts. 

d. Compaction of hydraulic fills. Dike-land 
hydraulic fills can be compacted as they are placed by 
use of- 

(1) Driving track-type tractors back and 
forth across the saturated fill. (Relative densities of 70 to 
80 percent can be obtained in this manner for 
cohesionless materials.) 

(2) Other methods such as vibratory 
rollers, vibroflotation, terraprobing, and compaction piles 
(chap 16). Below water, hydraulic fills can be compacted 
by use of terraprobing, compaction piles, and blasting. 

e. Underwater hydraulic fills. For structural fill 
placed on a dredged bottom, remove the fines dispersed 
in dredging by a final sweeping operation, preferably with 
suction dredges, before placing the fill. To prevent 
extremely flat slopes at the edge of a fill, avoid excessive 
turbulence during dumping of the fill material by placing 
with clamshell or by shoving off the sides of deck barges. 
To obtain relatively steep slopes in underwater fill, use 
mixed sand and gravel. With borrow containing about 
equal amounts of sand and gravel, underwater slopes as 
steep as 1V on 2H may be achieved by careful 
placement. Uncontrolled bottom dumping from barges 
through great depths of water will spread the fill over a 
wide area. To confine such fill, provide berms or dikes of 
the coarsest available material or stone on the fill 
perimeter. 



15-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 16 
STABILIZATION OF SUBGRADE SOILS 



16-1. General. 

a. The applicability and essential features of 
foundation soil tr eatments are summarized in tables 16-1 
and 16-2 and in lfigure 16-1.1 The depth of stabilization 
generally must be sufficient to absorb most of the 
foundation pressure bulb. 

b. The relative benefits of vibrocompaction, 
vibrodisplacement compaction, and precompression 
increase as load intensity decreases and size of loaded 
area increases. Soft, cohesive soils treated in place are 
generally suitable only for low-intensity loadings. Soil 
stabilization of wet, soft soils may be accomplished by 
addition of lime; grout to control water flow into 
excavations to reduce lateral support requirements or to 
reduce liquefaction or settlement caused by adjacent pile 
driving; seepage control by electroosmosis; and 
temporary stabilization by freezing. The range of soil 
grain sizes for which each stabilization method is most 
applicable is shown in lfigure 16T1 

16-2. Vibrocompaction. Vibrocompaction methods 
(blasting, terraprobe, and vibratory rollers) can be u sed 
for ra pid densification of saturated cohesionless soils Kfigl 
1 16-11) . The ranges of grain-size distributions suitable for 
treatment by vibrocompacti on, as well as vibroflotation, 
are shown in Ifigure 16-2. | The effectiveness of these 
methods is greatly reduced if the percent finer than the 
No. 200 sieve exceeds about 20 percent or if more than 
about 5 percent is finer than 0.002 millimeter, primarily 
because the hydraulic conductivity of such materials is 
too low to prevent rapid drainage following liquefaction. 
The usefulness of these methods in partly saturated 
sands is limited, because the lack of an increase of pore 
water pressure impedes liquefaction. Lack of complete 
saturation is less of a restriction to use of blasting 
because the high-intensity shock wave accompanying 
detonation displaces soil, leaving depressions that later 
can be backfilled. 

a. Blasting. 

(1) Theoretical design procedures for 
densification by blasting are not available and continuous 
on-site supervision by experienced engineers having 
authority to modify procedures as required is esential if 
this treatment method is used. A surface heave of about 
6 inches will be observed for proper charge sizes and 
placement depths. Surface cratering should be avoided. 
Charge masses of less than 4 to more than 60 pounds 



have been used. The effective radius of influence for 
charges using (M = lb) 60 percent dynamite is as follows: 

R = 3M 1/3 (feet) (16-1) 

Charge spacings of 10 to 25 feet are typical. The center 
of charges should be located at a depth of about two- 
thirds the thickness of the layer to be densified, and 
three to five successive detonations of several spaced 
charges each are likely to be more effective than a single 
large blast. Little densification is likely to result above 
about a 3-foot depth, and loosened material may remain 
around blast points. Firing patterns should be 
established to avoid the "boxing in" of pore water. Free- 
water escape on at least two sides is desirable. 

(2) If blasting is used in partly saturated 
sands or loess, preflooding of the site is desirable. In 
one technique, blast holes about 3 to 3/2 inches in 
diameter are drilled to the desired depth of treatment, 
then small charges connected by prima cord, or simply 
the prima cord alone, are strung the full depth of the 
hole. Each hole is detonated in succession, and the 
resulting large diameter holes formed by lateral 
displacement are backfilled. A sluiced-in cohesionless 
backfill will density under the action of vibrations from 
subsequent blasts. Finer grained backfills can be 
densified by tamping. 

b. Vibratingprobe (terraprobe). 

(1) A 30-inch-outside-diameter, open- 
ended pipe pile with 3/, -inch wall thickness is suspended 
from a vibratory pile driver operating at 15 Hz. A probe 
length 10 to 15 feet greater than the soil depth to be 
stabilized is used. Vibrations of 7%- to 1-inch amplitude 
are in a vertical mode. Probes are made at spacings of 
3 to 10 feet. After sinkage to the desired depth, the 
probe is held for 30 to 60 seconds before extraction. 
The total time required per probe is typically 21/2 to 4 
minutes. Effective treatment has been accomplished at 
depths of 12 to 60 feet. Areas in the range of 450 to 700 
square yards may be treated per machine per 8-hour 
shift. 

(2) Test sections about 30 to 60 feet on a 
side are desirable to evaluate the effectiveness and 
required probe spacing. The grain-size range of treat ed 
soil should fall within limits shown in figure 16-2. I A 
square pattern is often used, with a fifth probe at the 
center of each square giving more effective increased 
densification than a reduced spacing. Saturated soil 



16-1 



TM 5-818-1 / AFM 88-3, Chap. 7 









HOST SUITABLE 


MAXIMUM 

EPTECTIVE 

TREATMENT 

DEPTH 


ECONOMICAL 


SPECIAL 


SPECIAL 


PROPERTIES OF 


SPECIAL 


RELATIVE 




METHOD 


PRINCIPLE 


SOIL CONDITIONS/ 


size or 


MATERIALS 


EQUIPKBHT 


TREATED 


ADVANTAGES 


COSTS 








TYPES 


TREATED AREA 


REQUIRED 


REQUIRED 


MATERIAL 


AMD LIMITATIONS 


(197ft) 




•LASTING 


Shock wtvti and 


Saturated, claan 


60 ft 


Sam 11 areas 


Explosives, 


Jetting or 


Can obtain 


Rapid, inexpen- 


Low 






vibrations cause 


aandai partly 




can be 


backfill to 


drilling 


relative densities 


sive, can treat 


($0.50 






liquefaction, dia- 


saturated sands 




treated 


plug drill 


machine 


to 70~90%i may get 


small areasi 


11. 00 






placement, remold- 


and ailta aftar 




economically 


holes 




variable density 


variable proper- 


per cu yd 






ing 


flooding 












ties, no improve- 
ment near surface, 




8 


















dangerous 




TERRAPROBE 


Densification by 


saturated or dry 


60 ft 


>1200 yd 2 


Hone 


Vibratory 


Can obtain 


Rapid, simple. 


Moderate 


»* 




vibration) llque- 


claan aand 


(Ineffective 






pile driver 


Relative Densities 


good underwater; 


$1.50- 


5 




f act ion induced 




above 12 ft 






and 750 mm 


of 80% or more 


soft under layers 


$3.35 


S 




settlement under 




depth) 






die open 




may damp vibra- 


per cu yd 


§ 




overburden 










steel pipe 




tions, difficult 
to pane t rata. 


$2.00/cu yd 
average 


m 


















stiff overlayers. 




> 


















not good in 
partly satursted 
soils 




VIBRATORY 


Densification by 


Cohasionlaas 


6-10 ft 


Any sise 


None 


Vibratory 


Can obtain very 


Best method for 


Low 




ROLLERS 


vibration t lique- 
faction induced 
settlement under 
rollar weight 


soils 








rollar 


high relative 
densities i upper 
few decimeters 
not denalfiad 


thin layers or 
lifts 






COMPACTION 


Densification by 


Loos* sandy 


60 ft 


Snail to 


Pile Material 


Pile driver 


Can obtain high 


Useful in soils 


High 




PIUS 


displacement of 
pila volume and 
by vibration 
during driving 


soils i partly 
sat ura tad clayey 
soils i loess 




Moderate 


(often aand 
or soil plus 
cement 
Mixture) 




densities, good 
uniformity 


with fines, uni- 
form compaction, 
easy to chock 
results; slow, 
limited improve- 




s 


















ment in upper 




M 


















l-2l 




HEAVY TAMPING 


Repeated appli- 


Cohesion less 


50-60 ft 


>A0OO yd 2 


Hone 


Tamper of 


Can obtain high 


Simple, rapid, 


<Vibro- 





(Dynamic 


cation of high 


soils beat. 








10-40 tons 


relative densities, 


suitable for soma 


flotation 


I 


Consolidation) 


intanaity impacts 
at aurfaca 


other types can 
also be improved 








high capacity 
crane 


reasonable 
uniformity 


molls with fines j 
usable above and 
below water i 
requires control, 
must be away from 




M 

o 

4 

■ 

5 


















existing structures 




viworu>- 


Oanaification by 


Coheelonlese 


90 ft 


>1200 yd 2 


Granular 


vi&roflot. 


Can obtain high 


useful in satu- 


$10.00- 


TRTION 


vibration and 


soils with less 






backfill 


crane 


relative dammit lea, 


rated and partly 


$25.00/yd 






compaction of 


than 20% fines 










good uniformity 


saturated soils. 


$1.00/ en yd 






backfill material 














uniformity 


May be about 
half cKxapec- 
tion pile* or 
concrete piles 



(Continued) 



(Sheet 1 of 5) 



Table 16-1. Stabilization of Soils for Foundations of Structures 

16-2 



TM 5-818-1 / AFM 88-3, Chap. 7 









MOST SUITABLE 


MAXIMUM 

EFFECTIVE 

TREATMENT 

DEPTH 


ECONOMICAL 


SPECIAL 


SPECIAL 


PROPERTIES OP 


SPECIAL 


RELATIVE 




HBTHOO 


PRINCIPLE 


SOIL CONDITIONS/ 


size or 


MATERIALS 


EQUIPMENT 


TREATED 


ADVANTAGES 


COSTS 








TYPES 


TREATED AREA 


REQUIRED 


REQUIRED 


MATERIAL 


AMD LIMITATIONS 


(1976) 




PARTICULATE 


Penetration grout- 


Medium to coarse 


Unlimited 


Small 


Grout, water 


Misers, tanks, 


Impervious, high 


Low cost grouts, 


Lowest of 




GROUTING 


lng-flll «oll 
ports with soil, 
cement, and/or 
clay 


sand and gravel 








puaps, hoses 


strength with 
cement grout, 
eliminate lique- 
faction danger 


high strength; 
limited to 
coarse-grained 
soils, hard to 
evaluate 


ths grout 
systems 


CHEMICAL 


Solutions of two 


Medium silts and 


Ununited 


Small 


Grout, water 


Misers, tanks. 


Impervious, low 


Low viscosity. 


High to 




grouting 


or more enemies Is 
react in soil 
pores to for* a 
9*1 or a solid 


coarser 








pumps, hoses 


to high strength, 
eliminate lique- 
faction danger 


controllable gel 
time, good water 
■hut-off j high 
cost, hard to 


very high 
$30/m* min.- 
SSO/a'typ- 
icsl 


s 

M 




precipitate 














evaluate 




PRESSURE 


Lias slurry 


Expansive clays 


Unllnited, but 


Small 


Lime, water. 


Slurry tanks. 


Lime encapsulated 


Rapid and eco- 


$2.50 to 


M 


INJECTED LIMB 


Injected to 




2- 3 a usual 




surfactant 


agitators , 


sones formed by 


nomical treat- 


93.00/a of 


a 




shallow depths 










injection 


channels resulting 


ment for 


ground 


i 




under high 












from cracks, root 


foundation soils 


surface 


M 




prsssura 












holes, hydraulic 


under light 


ares 
















fracture 


structures 




DISPLACEMENT 


Highly viscous 


Soft, fine- 


Unlimited, but 


Small 


Soil, cement. 


Batching 


Grout bulbs within 


Good for correc- 


Low mate- 




GROUT 


grout acts as 


grained soilsi 


a few a usual 




water 


equipment , 


compressed soil 


tion of diffe- 


rial, high 






radial hydrau- 


foundation soils 








high pressure 


matrix 


rential settle- 


injection 






lic jack whan 


with large voids 








pumps, hoses 




ments, filling 








pumped In under 


or cavities 












large voids; 








high pr assure 














careful control 
required 




EUKTnoKimrric 


Stabilising 


Saturated silts. 


Unknown 


Small 


Chemical 


DC power 


Incressed strength. 


Existing soil end 


Expensive 




INJSCTIOM 


chemicals moved 
into soil by 
elect ro-ossos 1 s 


silty clays 






stabiliser 


supply, 
anodes, 
cathodes 


reduced compres- 
aibility 


structures not 
subjected to high 
pressures; no 
good in soil with 
high conductivity 





(Continued) 



(Sheet 2 of 5) 



Table 16-1. Stabilization of Soils for Foundations of Structures-Continued 



16-3 



TM 5-818-1 / AFM 88-3, Chap. 7 









HOST SUITABLE 


HA XI HUH 


ECONOMICAL 


SPECIAL 


SPECIAL 


PROPERTIES OP 


SPECIAL 


RELATIVE 




HCTHOD 


PRINCIPLE 


SOIL CONDITIONS/ 


EFFECTIVE 

TREATMENT 
DEPTH 


SIZE or 


MATERIALS 


EQUIPMENT 


TREATED 


ADVANTAGES 


COSTS 








TYPES 


TKEATED AREA 


REQUIRED 


REQUIRED 


MATERIAL 


AMD LIMITATIONS 


(1976) 




PRELOADING 


Load is appliad 


Normally consoli- 




>1000 m 1 


Earth fill or 


Earth moving 


Reduced water 


Easy, theory well 


Low 






sufficiently in 


dated soft clays. 






other material 


equipment i 


content and void 


developed, uni- 


(Moderate 






advance of con- 


silts, organic 






for loading 


large water 


ratio, increased 


formity; requires 


if verti- 






struction so that 


deposits, com- 






the site; sand 


tanks or 


strength 


long tie* (sand 


cal drains 






compression of 


pleted sanitary 






or gravel for 


vacuus 




drains or vlcks 


are re- 






soft soils is 


landfills 






drainage 


drainage 




can be used to 


quired) 






completed prior 








blanket 


systems so— 




reduce consolida- 








to development 










times used; 




tion time) 








of the site 










settlement 

markers, 

plesometers 










SURCHARGE 


Fill in excess 


Normally consoli- 




>1000 m* 


Earth fill or 


Earth moving 


Reduced water 


Paster than pre- 


Moderate. 




FILLS 


of that required 


dated soft clays. 






other material 


equipment ; 


content, void 


loading without 


Sand 






permanently is 


silts, organic 






for loading 


settlement 


ratio and cam 


surcharge, theory 


drains 






appliad to 


deposits, com- 






the site j sand 


markers. 


pressibillty; 


well developed; 


cost $3.30- 






achieve a given 


pleted sanitary 






or gravel for 


pieaometers 


increased strength 


extra material 


$6.60/m 


s 




amount of 


landfills 






drainage 






handling; can use 




M 




settlement in a 
shorter time; 
excess fill 








blanket 






sand drains or 
wicks 




s 

0- 




than removed 


















DYNAMIC 


High energy Im- 


Partly saturated 


20 m 


>15000- 
30000 m 


None 


Tamper of 


Reduced water 


Paster than pre- 


< preload 




CONSOLIDATION 


pacts compress and 


fine grained 






10-40 tons. 


content, void 


loading, economi- 


fills with 






dissolve gas In 


soils, quaternary 








high capacity 


ratio and com- 


cal on large 


sand 






pores to give 


clays with 1-4 








crane 


pressibility; 


areas; uncertain 


drains 






immediate settle- 


gas in micro 










increased strength 


mechanism in 








ment; Increased 


bubbles 












clays, less uni- 








pore pressure 














formity than 








gives subsequent 














preloading 








drainage 


















ELECTRO- 


DC currant 


Normally consoli- 


10-20 m 


Small 


Anodes 


DC power 


Reduced water 


No fill loading 


High 




OSMOSIS 


causes water flow 


dated silts and 






(usually re- 


supply, 


content and com 


required, can use 








from anode towards 


silty clays 






bars or 


wiring , 


possibility. 


in confined area. 








cathode where it 








aluminum). 


metering 


increased strength. 


relatively fasti 








is removed 








cathodes 
(well points 
or rebars) 


systems 


electrochemical 
hardening 


non-uniform pro- 
perties between 
electrodes, no 
good in highly 
conductive soils 





(Continued) 



(Sheet 3 of 5) 



Table 16-1. Stabilization of Soils for Foundations of Structures-Continued 



16-4 



TM 5-818-1 / AFM 88-3, Chap. 7 





METHOD 


PRINCIPLE 


MOST SUITABLE 

SOIL CONDITIONS/ 

TYPES 


MAXIMUM 

EFFECTIVE 

TREATMENT 

DEPTH 


ECONOMICAL 

SIZE OF 

TREATED AREA 


SPECIAL 
MATERIALS 
REQUIRED 


SPECIAL 
EQUIPMENT 
REQUIRED 


PROPERTIES OF 
TREATED 
MATERIAL 


SPECIAL 
ADVANTAGES 
AND LIMITATIONS 


RELATIVE 
COSTS 
(1976) 




HIX-IN-PLACE 
PILES AND HALLS 


Line, ctMnt, or 
asphalt introduced 
through rotating 
auger or spacial 
in-placa nixar 


All aoft or loose 
inorganic soils 


>20 * 


Snail 


Ceeant, line, 
asphalt, or 
chemical 
stabiliser 


Drill rig, 
rotary cut- 
ting and 
nixing head, 
additive 
proportion- 
ing aquip- 
satnt 


Solidified aoil 
pilaa or walls of 
relatively high 
strength 


Uses native soil, 
reduced lateral 
support require- 
aanta during 
excavation) 
difficult quality 
control 


Moderate 
to high 




i 


STRIPS AND 
HBMMAMBS 


Horizontal tensile 
■trips or ■■■ 
branee buried in 
•oil under foot- 
inga 


All 


A few a 


Snail 


Metal or 
plastic strips, 
polyethylene , 
polypropylene 
or polyester 
fabrics 


Excavating, 
earth handling, 
and oonpaction 
equipment 


Increaaed bearing 
capacity, reduced 
deformations 


Increaaed allow- 
able bearing 
pressure) 
requires over- 
escavation for 
footings 


Low to 
Moderate 




VIBRO-REFLACE- 
MBIT STONE 
COLUMNS 


Hoi* jatted into 
aoft, fina-gzainad 
soil and back- 
filled with 
danaely coapactad 
gravel 


Soft clays and 
alluvial deposits 


20 ■ 


>1500 m* 


Gravel or 
crushed rock 
backfill 


Vibroflot, 
crane or 
Vlbrocat, 
water 


Increased bearing 
capacity, reduced 

settlements 


Faster than pre- 
compression, 
avoids dewator- 
ing required for 
remove and 
replace t limited 
bearing capacity 


Moderate 

to high 
-%30/mt 
>pile 
penetra- 
tion 


1 

t 


MUTING 


Drying at low 
tempera turea ; 
alteration of 
claya at inter- 
mediate teapsr- 
aturaa (400- 
600 «C)i 

fuaion at high 
temparaturee 
(>1000»C) 


Pina-grainad 
•oils, especially 
partly saturated 
clays and silts, 
loaas 


15 a 


SJMll 


Fuel 


Fuel tanks, 

burners, 

blowers 


Reduced water 
control, plas- 
ticity, water 
sensitivity! 
Increased strength 


Can obtain lr- 
ravsrslbla 
improvements in 
properties) can 
introduce etabi- 
liaars with hot 
gases. Experi- 
mental 


High 






PRESSING 


Freese aoft, wat 
ground to ln- 
craaaa its 
strength and 

Stiff MSB 


All soils 


Several si 


Sse.ll 


Refrigerant 


Refrigeration 
systen 


Increased strength 
and stiffness i 
reduced perme- 
ability 


No good in 
flowing ground 
water, temporary 


High 



(Continued) 



(Sheet U of 5) 



Table 16-1. Stabilization of Soils for Foundations of Structures-Continued 



16-5 



TM 5-818-1 / AFM 88-3, Chap. 7 









HOST SUITABLE 


MAXIMUM 

EFFECTIVE 

TREATMENT 

DEPTH 


ECONOMICAL 


SPECIAL 


SPECIAL 


PROPERTIES OF 


SPECIAL 


RELATIVE 




METHOD 


PRINCIPLE 


SOIL CONDITIONS/ 


SIZE OF 


MATERIALS 


EQUIPMENT 


TREATED 


ADVANTAGES 


COSTS 








TYPES 


TREATED AREA 


REQUIRED 


REQUIRED 


MATERIAL 


AND LIMITATIONS 


(1976) 




REMOVE AND 


Foundation soil 


Inorganic soils 


10 m ? 


Small 


None, unless 


Excavating 


Increased strength 


Uniform, con- 


High 




REPLACE 


excavated , im- 








admixture 


and compac- 


and stiffness. 


trolled founda- 






(with or without 


proved by drying 








stabilizers 


tion equip- 


reduced compressi- 


tion soil when 






admixtures) 


or admixture, and 
r» compacted 








used 


ment, 

dewa taring 
system 


bility 


replaced; may 
require large 
area dewatering 




HOI STUPE 


Water access to 


Expansive soils 


5 m 


Small 


Membranes , 


Excavating 


Initial natural or 


Best for small 


Low to 


( 


BARRIERS 


foundation soils 
is prevented 








gravel, lime, 
or asphalt 


and trench- 
ing equip- 
ment, 

compaction 
equipment 


as-compacted 
properties 

retained 


structures; amy 
not be 100% 
effective 


■ode rate 


PREHETTIMG 


Soil is brought 


Expansive soils 


2-3 m 


Small 


Hater 


None 


Decreased swell- 


Low cost, best 


Low 


M 




to estimated 












ing potential 


for small, light 






final water 














structures, may 








content prior to 
construction 














still get shrink- 
ing and swelling 




STRUCTURAL PILLS 


Structural fill 


Use over soft 




Small 


Sand, gravel, 


Compaction 


Soft subgrade 


High strength. 


Moderate 




(with or without 


distributes 


clays or organic 






flyash. 


equipment 


protected by 


good load dis- 


to high 




Admixtures) 


loads to under- 
lying soft soils 


soils, marsh 
deposits 






bottom ash, 
slag, expand- 
ed aggregate, 
clam shell or 
oyster shell, 
incinerator 
ash 




structural load- 
bearing fill 


tribution to 
underlying soft 
soils 


(512/m 3 ) 



(Sheet 5 of 5) 



U. S. Army Corps of Engineers 



Table 16-1. Stabilization of Soils for Foundations of Structures-Continued 



16-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



CATEGORY OP 
STRUCTURE 


STRUCTURE 


PERMISSIBLE 
SETTLEMENT 


LOAD INTENSITY/ 
USUAL BEARING PRESSURE 
REQUIRED (tsf) 


Probability of Advantageous Use of Soil Improvement 
Techniques 


LOOSE COHESION- 
LESS SOILS 


SOFT ALLUVIAL 
DEPOSITS 


OLD, INORGANIC PILLS 


OFFICE/ 
APARTMENT 

FRAME OR LOAD- 
BEARING 
CONSTUCTIOM 


Nigh rite: Mora 
than 6 stories 


Small 
<2S-S0 mm 


High (3+1) 


High 


Unlikely 


Low 


Medium rise: 
3-6 stories 


Small 
<25-S0 mm 


Moderate (2) 


High 


Lov 


Good 


Low rise: 
1-3 stories 


Small 
<2S-S0 mm 


Low (1-2) 


High 


Good 


High 


INDUSTRIAL 


Large span w/heavy 
machines, cranesi 
process plants: 
power plants 


Small <<2S-S0 mm) 

Differential settlement 

Critical 


Variable/high local 
concentrations to >U 


High 


Unlikely 


Low 


Framed warehouses 
and factories 


Moderate 


Lov (1-2) 


High 


Good 


High 


Covered storage, 
stor. rack systems, 
production areas 


Low to moderate 


Low (<2) 


High 


Good 


High 


OTHERS 


Mater and waste 
water treatment 
plants 


Moderate 

Differential settlement 

Important 


Low/<150 (<1.S) 


High, if required 
at all 


High 


High 


Storage tanks 


Moderate to high, 
but differential, 
may be critical 


High/up to 300 (!) 


High, if required 
at all 


High 


High 


Open storage areas 


High 


High/up to 300 (3) 


High, if required 
at all 


High 


Ugh 


Embankments and 
abutments 


Moderate to high 


High/up to 200 (2) 


High, if required 
at- all 


High 


High 



U. S. Army Corps of Engineers 

Table 16-2. Applicability of Foundation Soil Improvement for Different Structures and Soil Types (for efficient use of Shallow Foundations) 

16-7 



TM 5-818-1 / AFM 88-3, Chap. 7 



Grave\ 



Sand 



Silt 



Cloy 



ViDro- compoction 



Vidro-displacement Compaction 



Particulate Grout ~] 



3 



Chemical Grout 



£ 



Displacement Grout 



I 



3 



Preloading 



Dynamic Consolidation (heavy tamping) 



£ 



Electro-osmosis 



Reinforcement 



Thermal Treatment 



Remove and Reploce 



1 Presetting 



10 



1.0 



0.1 0.01 

Particle Size - mm 



0.001 



OOOOI 



(Courtesy of J. K. Mitchell, "Innovations in Ground Stabilization, " Chicago Soil Mechanics Lecture 
Series, Innovations in Foundation Construction, Illinois Section, 1972. Reprinted by permission of The 
American Society of Civil Engineers, New York.) 

Figure 16-1. Applicable grain-size ranges for different stabilization methods. 

16-8 



TM 5-818-1 / AFM 88-3, Chap. 7 



^80 
I 

£ 40 



c 

O 

V. 



20 



Grovel 




< 


Sand 




i 




Silt 




Clay 


Tl ' 


^//KXXX 


p<XX/ 


































Most Desirable Size Range 




V 

\ 




v^OO 








(/A \ 


(Vibroflototion) 




\ 
> 


























\ 

N 
\ 


























\ 

\ 






















V77? 


i 


i c 


? /.0 0.5 


0.2 0.1 0.05 0.02 0.01 0.005 


0001 



Particle Size - mm 

(Courtesy of J. K. Mitchell, "Innovations in Ground Stabilization, " Chicago Soil Mechanics Lecture 
Series, Innovations in Foundation Construction, Illinois Section, 1972. Reprinted by permission of The 
American Society of Civil Engineers, New York.) 

Figure 16-2. Range of particle-size distributions suitable for densification by vibrocompaction. 



16-9 



TM 5-818-1 / AFM 88-3, Chap. 7 



conditions are necessary. Underlying soft clay layers 
may dampen vibrations. 

c. Vibratory rollers. Where cohesionless 
deposits are of limited thickness, e.g., less than 6 feet, or 
where cohesionless fills are being placed, vibratory 
rollers are likely to be the best and most economical 
means for achieving high density and strength. Use with 
flooding where a source of water is available. The 
effective depth of densificatio n may be 6 feet or more for 
the heaviest vibratory rollers |fig 16-3a| ). For a fill placed 
in suc cessive lifts, a density-depth distribution similar to 
that in I figure 16-"3B results. It is essential that the lift 
thickness, soil type, and roller type be matched. Properly 
matched systems can yield compacted layers at a 
relative density of 85 to 90 percent or more. 

16-3. Vibrodisplacement compaction. The methods 
in this group are similar to those described in the 
preceding section except that the vibrations are 
supplemented by active displacement of the soil and, in 
the case of vibroflotation and compaction piles, by 
backfilling the zones from which the soil has been 
displaced. 

a. Compaction piles. Partly saturated or freely 
draining soils can be effectively densified and 
strengthened by this method, which involves driving 
displacement piles at close spacings, usually 3 to 6 feet 
on centers. One effective procedure is to cap 
temporarily the end of a pipe pile, e.g., by a detachable 
plate, and drive it to the desired depth, which may be up 
to 60 feet. Either an impact hammer or a vibratory driver 
can be used. Sand or other backfill material is 
introduced in lifts with each lift compacted concurrently 
with withdrawal of the pipe pile. In this way, not only is 
the backfill compacted, but the compacted column has 
also expanded laterally below the pipe tip forming a 
caisson pile. 

b. Heavy tamping (dynamic consolidation). 

(1) Repeated impacts of a very heavy 
weight (up to 80 kips) dropped from a height of 50 to 130 
feet are applied to points spaced 15 to 30 feet apart over 
the area to be densified. In the case of cohesionless 
soils, the impact energy causes liquefaction followed by 
settlement as water drains. Radial fissures that form 
around the impact points, in some soils, facilitate 
drainage. The method has been used successfully to 
treat soils both above and below the water table. 

(2) The product of tamper mass and 
height of fall should exceed the square of the thickness 
of layer to be densified. A total tamping energy of 2 to 3 
blows per square yard is used. Increased efficiency is 
obtained if the impact velocity exceeds the wave velocity 
in the liquefying soil. One crane and tamper can treat 
from 350 to 750 square yards per day. Economical use 
of the method in sands requires a minimum treatment 
area of 7500 square yards. Relative densities of 70 to 90 



percent are obtained. Bearing capacity increases of 200 
to 400 percent are usual for sands and marls, with a 
corresponding increase in deformation modulus. The 
cost is reported as low as one-fourth to one-third that of 
vibroflotation. 

(3) Because of the high-amplitude, low- 
frequency vibrations (2-12 Hz), minimum distances 
should be maintained from adjacent facilities as follows: 
Piles or bridge abutment 1 5-20 feet 

Liquid storage tanks 30 feet 

Reinforced concrete buildings 50 feet 
Dwellings 100 feet 
Computers (not isolated) 300 feet 

c. Vibroflotation. 

(1) A cylindrical penetrator about 15 
inches in diameter and 6 feet long, called a vibroflot, is 
attached to an adapter section containing lead wires and 
hoses. The whole assembly is handled by a crane. A 
rotating eccentric weight inside the vibroflot develops a 
horizontal centrifugal force of about 10 tons at 1800 
revolutions per minute. Total weight is about 2 tons. 

(2) To sink the vibroflot to the desired 
treatment depth, a water jet at the tip is opened and acts 
in conjunction with the vibrations so that a hole can be 
advanced at a rate of about 3.6 feet per minute; then the 
bottom jet is closed, and the vibroflot is withdrawn at a 
rate of about 0.1 foot per minute. Newer, heavier 
vibroflots operating at 100 horsepower can be withdrawn 
at twice this rate and have a greater effective penetration 
depth. Concurrently, a cohesionless sand or gravel 
backfill is dumped in from the ground surface and 
densified. Backfill consumption is at a rate of about 0.7 
to 2 cubic yards per square yard of surface. In partly 
saturated sands, water jets at the top of the vibroflot can 
be opened to facilitate liquefaction and densification of 
the surrounding ground. Liquefaction occurs to a radial 
distance of 1 to 2 feet from the surface of the vibroflot. 
Most vibroflotation applications have been to depths less 
than 60 feet, although depths of 90 feet have been 
attained successfully. 

(3) A relationship between probable 
relative den sity and vibroflot hold spacings is given in 

I figure 16^4] Newer vib roflots result in greater relative 
densities. I Figure 16-5 shows relationships between 
allowable bearing pressure to limit settlements to 1 inch 
and vibroflot spacing. Allowable pressures for 
"essentially cohesionless fills" are less than for clean 
sand deposits, because such fills invariably contain 
some fines and are harder to density. 

(4) Continuous square or triangular 
patterns are often used over a building site. 
Alternatively, it may be desired to improve the soil only at 
the locations of individual spread footings. Patterns and 
spacings required for an allowable pressure of 3 tons p er 
square foot and square footings are given in | table ^"^ 



16-10 



TM 5-818-1 / AFM 88-3, Chap. 7 



DRY DENSITY, PCF 

100 105 




90 



DRY DENSITY, PCF 
100 105 





















NOTE: 2-FT LIFT HEIGHT \ 
5 ROLLER PASSES. \ 
I \ 


) 



60 ao 

RELATIVE DENSITY, PERCENT 



100 



60 80 

RELATIVE DENSITY, PERCENT 



a. DENSITY VERSUS DEPTH FOR DIFFERENT 
NUMBERS OF ROLLER PASSES 



b. DENSITY VERSUS DEPTH RELATIONSHIP FOR A 
SERIES OF 2-FT LIFTS 



Figure 16-3. Sand densification using vibratory rollers. 



16-11 



TM 5-818-1 / AFM 88-3, Chap. 7 



16-4. Grouting and injection. Grouting is a high-cost 
soil stabilization method that can be used where there is 
sufficient confinement to permit required injection 
pressures. It is usually limited to zones of relatively small 
volume and to special problems. Some of the more 
important applications are control of groundwater during 
construction; void filling to prevent excessive settlement; 
strengthening adjacent foundation soils to protect against 
damage during excavation, pile driving, etc.; soil 
strengthening to reduce lateral support requirements; 
stabilization of loose sands against liquefaction; 
foundation underpinning; reduction of machine 
foundation vibrations; and filling solution voids in 
calcareous materials. 

a. Grout types and groutability. Grouts can be 
classified as particulate or chemical. Portland cement is 
the most widely used particulate grouting material. 
Grouts composed of cement and clary are also widely 
used, and lime-slurry injection is finding increasing 
application. Because of the silt-size particles in these 
materials, they cannot be injected into the pores of soils 
finer than medium to coarse sand. For successful 
grouting of soils, use the following guide 



(Dl5)soil 

(L'ssJgrout 



>25 



Type I portland cement, Typeportland cement, and Type 
III portland cement, Type III portland cement, and 
processed bentonite cannot be used to penetrate soils 
finer than 30, 40, and 60 mesh sieve sizes, respectively. 
Different types of grouts may be combined to both 
coarse- and fine-grained soils. 

b. Cement and soil-cement grouting. See TM 
5-818-6/AFM 88-32 for discussion of planning and 
implementation of foundation grouting with cement and 
soil-cement. 

c. Chemical grouting. To penetrate the voids 
of finer soils, chemical grout must be used. The most 
common classes of chemical grouts in current use are 
silicates, resins, lignins, and acrylamides. The viscosity 
of the chemical-water solution is the major factor 
controlling groutability. The particle-size ranges over 



100 



o 

DC 



Ol 



CO 

z 



ID 
> 



oc 



03 
< 

CO 

O 
cc 
o. 



90 



80 



70 



60 



50 






1 5 10 

SPACING OF CENTERS OF VIBRATION, FT 



U. S. Army Corps of Engineers 

Figure 16-4. Relative density as a function of vibroflot hole spacings. 



16-12 



TM 5-818-1 / AFM 88-3, Chap. 7 




5 10 

SPACINGS OF CENTERS OF VIBRATION, FT 



U. S. Army Corps of Engineers 

Figure 16-5. Allowable bearing pressure on cohesionless soil layers stabilized by vibroflotation. 



Table 16-3. Vibroflotation Patterns for Isolated Footings for an Allowable Bearing Pressure. 



Square Footing 


Vibroflotation 


Center to Center 




Size, ft 


Points 


Spacing, 


ft 


Pattern 


4.0 


1 




4.5-5.5 


2 


6.0 




Line 


6-7 


3 


7.5 




Triangle 


7.5-9.5 


4 


6.0 




Square 


10-12 


5 


7.5 




Square + 1 

@ Center 



U. S. Army Corps of Engineers 



16-13 



TM 5-818-1 / AFM 88-3, Chap. 7 



which each of these grout types is effective is shown in 
I figure 16-61 

16-5. Precompression. 

a. Preloading. Earth fill or other material is 
placed over the site to be stabilized in amounts sufficient 
to produce a stress in the soft soil egual to that 
anticipated from the final structures. As the time 
required for consolidation of the soft soil may be long 
(months to years), varying directly as the square of the 
layer thickness and inversely as the hydraulic 
conductivity, preloading alone is likely to be suitable only 
for stabilizing thin layers and with a long period of time 
available prior to final development of the site. 

b. Surcharge fills. If the thickness of the fill 
placed for pre-loading is greater than that required to 
induce stresses corresponding to structure-induced 
stresses, the excess fill is termed a surcharge fill. 
Although the rate of consolidation is essentially 
independent of stress increase, the amount of 
consolidation varies approximately in proportion to the 
stress increase. It follows, therefore, that the preloading 
fill plus surcharge can cause a given amount of 
settlement in shorter time than can the preloading fill 
alone. Thus, through the use of surcharge fills, the time 
required for preloading can be reduced significantly. 

(1) The required surcharge and loading 
period can be determined using conventional theories of 
consolidation. Both primary consolidation and most of 
the secondary compression settlements can be taken 
out in advance by surcharge fills. Secondary 
compression settlements may be the major part of the 
total settlement of highly organic deposits or old sanitary 
landfill sites. 

(2) Because the degree of consolidation 
and applied stress vary with depth, it is necessary to 
determine if excess pore pressures will remain at any 
depth after surcharge removal. If so, further primary 
consolidation settlement under permanent loadings 
would occur. To avoid this occurrence, determine the 
duration of the surcharge loading required for points 
most distant from drainage boundaries. 

(3) The rate and amount of preload may 
be controlled by the strength of the underlying soft soil. 
Use berms to maintain foundation stability and place fill 
in stages to permit the soil to gain strength from 
consolidation. Predictions of the rates of consolidation 
strength and strength gain should be checked during fill 
placement by means of piezometers, borings, laboratory 
tests, and in situ strength tests. 

c. Vertical drains. 

(1) The required preloading time for most 
soft clay deposits more than about 5 to 10 feet thick will 
be large. The consolidation time may be reduced by 
providing a shorter drainage path by installing vertical 
sand drains. Sand drains are typically 10 to 15 inches 



in diameter and are installed at spacings of 5 to 15 feet. 
A sand blanket or a collector drain system is placed over 
the surface to facilitate drainage. Other types of drains 
available are special cardboard or combination plastic- 
cardboard drains. Provisions should be made to monitor 
pore pressures and settlements with time to determine 
when the desired degree of precompression has been 
obtained. 

(2) Both displacement and 

nondisplacement methods have been used for installing 
sand drains. Although driven, displacement drains are 
less expensive than augered or "bored" nondisplacement 
drains; they should not be used in sensitive deposits or in 
stratified soils that have higher hydraulic conductivity in 
the horizontal than in the vertical direction. Vertical 
drains are not needed in fibrous organic deposits 
because the hydraulic conductivity of these materials is 
high, but they may be required in underlying soft clays. 

d. Dynamic consolidation (heavy tamping). 
Densification by heavy tamping has also been reported 
as an effective means for improving silts and clays, with 
preconstruction settlements obtained about 2 to 3 times 
the predicted construction settlement. The time required 
for treatment is less than for surcharge loading with sand 
drains. The method is essentially the same as that used 
for cohesionless soils, except that more time is required. 
Several blows are applied at each location followed by a 
1- to 4-week rest period, then the process is repeated. 
Several cycles may be required. In each cycle the 
settlement is immediate, followed by drainage of pore 
water. Drainage is facilitated by the radial fissures that 
form around impact points and by the use of horizontal 
and peripheral drains. Because of the necessity for a 
time lapse between successive cycles of heavy tamping 
when treating silts and clays, a minimum treatment area 
of 18,000 to 35,000 square yards (4 to 8 acres) is 
necessary for economical use of the method. This 
method is presently considered experimental in 
saturated clays. 

e. Electroosmosis. Soil stabilization by 
electroosmosis may be effective and economical under 
the following conditions: (1) a saturated silt or silty clay 
soil, (2) a normally consolidated soil, and (3) a low pore 
water electrolyte concentration. Gas generation and 
drying and fissuring at the electrodes can impair the 
efficiency of the method and limit the magnitude of 
consolidation pressures that develop. Treatment results 
in nonuniform changes in properties between electrodes 
because the induced consolidation depends on the 
voltage, and the voltage varies between anode and 
cathode. Thus, reversal of electrode polarity may be 
desirable to achieve a more uniform stress condition. 
Electroosmosis may also be used to accelerate the 
consolidation under a preload or surcharge fill. The 
method is relatively expensive. 



16-14 



TM 5-818-1 / AFM 88-3, Chap. 7 



GRAVEL 



SAND 



TT 



SILT 



CLAY 



PORTLAND CEMENT 



SILICATES 



RESINS 



LIGNINS 



CONCENTRATION, %) 
I I I I I I l 
VISCOSITY, CPS (WATER=1) 



[COMPRESSIVE STRENGTH, KIP/FT 2 ] 

I* I I 



10) 1.5, (30) 15, [<36->360] 



(10) 



1.4, (30) 3.5, (40) 9, (50) 30 [>360] 



(10) 1.4,(30) 11, [<36] 



ACRYLAMIDES 



(10) 1.3, (30) 1.7, (50) 2.5 
[36-360] 



10 



1.0 



0.1 
SOIL PARTICLE SIZE, MM 



0.01 



0.001 



U. S. Army Corps of Engineers 

Figure 16-6. Soil particle sizes suitable for different grout types and several concentrations and viscosities shown. 



16-15 



TM 5-818-1 / AFM 88-3, Chap. 7 



16-6. Reinforcement. The supporting capacity of 
soft, compressible ground may be increased and 
settlement reduced through use of compression 
reinforcement in the direction parallel to the applied 
stress or tensile reinforcement in planes normal to the 
direction of applied stress. Commonly used 

compression reinforcement elements include mix-in- 
place piles and walls. Strips and membranes are used 
for tensile reinforcement, with the latter sometimes used 
to form a moisture barrier as well. 

a. Mix-in-place piles and walls. Several 
procedures are available, most of them patented or 
proprietary, which enable construction of soil-cement or 
soil-lime in situ. A special hollow rod with rotating vanes 
is augered into the ground to the desired depth. 
Simultaneously, the stabilizing admixture is introduced. 
The result is a pile of up to 2 feet in diameter. Cement, 
in amounts of 5 to 10 percent of the dry soil weight, is 
best for use in sandy soils. Compressive strengths in 
excess of 200 kips per square foot can be obtained in 
hese materials. Lime is effective in both expansive 
plastic clays and in saturated soft clay. Compressive 
strengths of about 20 to 40 kips per square foot are to be 
expected in these materials. If overlapping piles are 
formed, a mix-in-place wall results. 

b. Vibroreplacement stone columns. A 
vibroflot is used to make a cylindrical, vertical hole under 
its own weight by jetting to the desired depth. Then, 1/2- 
to 1- cubic yard coarse granular backfill, usually gravel or 
crushed rock 3/4 to 1 inch is dumped in, and the vibroflot 
is used to compact the gravel vertically and radially into 
the surrounding soft soil. The process of backfilling and 
compaction by vibration is continued until the densified 
stone column reaches the surface. 

c. Strips and membranes. 

(1) Low-cost, durable waterproof 
membranes, such as polyethylene, polypropolylene 
asphalt, and polyester fabric asphalt, have had 
application as moisture barriers. At the same time, these 
materials have sufficient tensile strength that when used 
in envelope construction, such as surrounding a well- 
compacted, fine-grained soil, the composite structure 
has a greater resistance to applied loads than 
conventional construction with granular materials. The 
reason is that any deformation of the enveloped soil layer 
causes tension in the membrance, which in turn 
produces additional confinement on the soil and thus 
increases its resist- ance to further deformation. 

(2) In the case of a granular soil where 
moisture infiltration is not likely to be detrimental to 
strength, horizontally bedded thin, flat metal or plastic 
strips can act as tensile reinforcing elements. 
Reinforced earth has been used mainly for earth 
retaining structures; however, the feasibility of using 
reinforced earth slabs to improve the bearing capacity of 
granular soil has been demonstrated. 



(3) Model tests have shown that the 
ultimate bearing capacity can be increased by a factor of 
2 to 4 for the same soil unreinforced. For these tests, 
the spacing between reinforcing layers was 0.3 times the 
footing width. Aggregate strip width was 42 percent of 
the length of strip footing. 

d. Thermal methods. Thermal methods of 
founda- tion soil stabilization, freezing or heating, are 
complex and their costs are high. 

(1) Artificial ground freezing. Frozen soil 
is far stronger and less pervious than unfrozen ground. 
Hence, artificial ground freezing has had application for 
temporary underpinning and excavation stabilization. 
More recent applications have been made to back- 
freezing soil around pile foundations in permafrost and 
maintenance of frozen soil under heated buildings on 
permafrost. Design involves two classes of problems; 
namely, the structural properties of the frozen ground to 
include the strength and the stress-strain-time behavior, 
and thermal considerations to include heat flow, transfer 
of water to ice, and design of the refrigeration system. 

(2) Heating. Heating fine-grained soils to 
moderate temperatures, e.g., 1000C+, can cause drying 
and accompanying strength increase if subsequent 
rewetting is prevented. Heating to higher temperatures 
can result in significant permanent property 
improvements, including decreases in water sensitivity, 
swelling, and compressibility; and increases in strength. 
Burning of liquid or gas fuels in boreholes or injection of 
hot air into 6- to 9-inch-diameter boreholes can produce 
4- to 7-foot-diameter strengthened zones, after 
continuous treatment for about 10 days. Dry or partly 
saturated weak clayey soils and loess are well suited for 
this type of treatment, which is presently regarded as 
experimental. 

16-7. Miscellaneous methods. 

a. Remove and replace. Removal of poor soil 
and replacement with the same soil treated by 
compaction, with or without admixtures, or by a higher 
quality material offer an excellent opportunity for 
producing high-strength, relatively incompressible, 
uniform foundation conditions. The cost of removal and 
replacement of thick deposits is high because of the 
need for excavation and materials handling, processing, 
and recompaction. Occasionally, an expensive 
dewatering system also may be required. Excluding 
highly organic soils, peats and sanitary landfills, virtually 
any inorganic soil can be processed and treated so as to 
form an acceptable structural fill material. 

b. Lime treatment. This treatment of plastic 
fine- 



16-16 



TM 5-818-1 / AFM 88-3, Chap. 7 



grained soils can produce high-strength, durable 
materials. Lime treatment levels of 3 to 8 percent by 
weight of dry soil are typical. 

c. Portland cement. With treatment levels of 3 
to 10 percent by dry weight, portland cement is 
particularly well suited for low-plasticity soils and sand 
soils. 

d. Stabilization using fills. 

(1 ) At sites underlain by soft, compressible 
soils and where filling is required or possible to establish 
the final ground elevation, load-bearing structural fills can 
be used to distribute the stresses from light structures. 
Compacted sands and gravels are well suited for this 
application as are also fly ash, bottom ash, slag, and 
various lightweight aggregates, such as expended shale, 
clam and oyster shell, and incinerator ash. Admixture 
stabilizers may be incorporated in these materials to 
increase their strength and stiffness. 



(2) Clam and oyster shells as a structural 
fill material over soft marsh deposits represent a new 
development. The large deposits of clam and oyster or 
reef shells that are available in the Gulf States coastal 
areas can be mined and tasportd short distances 
economically. Clam shells are 1/4, to 1/2 inch in 
diameter; whereas, oyster shells, which are coarser and 
more elongated, are 2 to 4 inches in size. When 
dumped over soft ground, the shells interlock; if there are 
finmes and water present, some cementation develops 
owing to the high calcium carbonate (>90 percent) 
content. In the loose state, the shell unit weight is about 
63 pounds per square foot; after construction, it is about 
95 pounds per square foot. Shell embenkments "float" 
over very soft ground; whereas, conventional fills would 
sink out of sight. About a 5-foot-thick layer is required to 
be placed in a single lift. The only compaction used is 
from the top of the lift, so the upper several inches are 
more tightly knit and denser than the rest of the layer. 



16-17 



TM 5-818-1 / AFM 88-3, Chap. 7 
CHAPTER 17 
DESIGN FOR EQUIPMENT VIBRATIONS AND SEISMIC LOADINGS 



17-1. Introduction. 

a. Vibrations caused by steady state or transient 
loads may cause settlement of soils, excessive motions 
of foundations or structures, or discomfort or distress to 
personnel. Some basic design factors for dynamic 
loading are treated in this section. Design of a 
foundation system incorporates the equipment loading, 
subsurface material properties, and geometrical 
proportion s in some ana lytical procedure. 

b. \ Figure 17-H shows some limiting values of 
vibration criteria for machines, structures, and personnel. 
On this diagram, vibration characteristics are described 
in terms of frequency and peak amplitudes of 
acceleration, velocity, or displacement. Values of 
frequency constitute the abscissa of the diagram and 
peak velocity is the ordinate. Values of peak 
displacement are read along one set of diagonal lines 
and labelled in displacement (inches), and peak 
acceleration values are read along the other set of 
diagonal lines and labelled in various amounts of g, the 
acceleration of gravity. The shaded zones in the upper 
right-hand corner indicate possible structural damage to 
walls by steady-state vibrations. For structural safety 
during blasting, limit peak velocity to 2.0 inches per 
second and peak acceleration to 0.1 Og for frequencies 
exceeding 3 cycles per second. These limits may 
occasionally have to be lowered to avoid being 
excessively annoying to people. 

c. For equipment vibrations, limiting criteria 
consist of a maximum velocity of 1 .0 inch per second up 
to a frequency of about 30 cycles per second and a peak 
acceleration of 0.1 5g above this frequency. However, 
this upper limit is for safety only, and specific criteria 
must be established for each installation. Usually, 
operating limits of equipment are based on velocity 
criteria; greater than 0.5 inch per second indicates 
extremely rough operation and machinery should be shut 
down; up to 0.10 inch per second occurs for smooth, 
well-balanced equipment; and less than 0.01 inch per 
second re presents very sm ooth operation. 

d. I Figure 17-11 also includes peak velocity 
criteria for reaction of personnel to steady-state 
vibrations. Peak velocities greater than 0.1 inch per 
second are "troublesome to persons," and peak 
velocities of 0.01 inch per second are just "barely 
noticeable to persons." It is significant that persons and 
machines respond to equivalent levels of vibration. 



Furthermore, persons may notice vibrations that are 
about 1/100 of the value related to safety of structures. 

17-2. Single degree of freedom, damped, forced 
systems. 

a. Vibrations of foundation-soil systems can 
adequately be represented by simple mass-spring- 
dashpot systems. The model for this simple system 
consists of a concentrated mass, m, supported by a 
linear elastic spring with a spring constant, k, and a 
viscous damping unit (dashpot) having a damping 
constant, c. The system is excited by an external force, 
e.g., Q = Q sin (cot), in which Q is the amplitude of the 
exciting force, co = 2jtf is the angular frequency (radians 
per second) with f the exciting frequency (cycles per 
second), and t is time in seconds. 

b. If the m odel is oriented as shown in the 
insert in Ifiqure 17-2(| a), motions will occur in the vertical 
or z direction only, and the system has one degree of 
freedom (one coordinate direction (z) is needed to 
describe the motion). The magnitude of dynamic vertical 
motion, A z , depends upon the magnitude of the external 
excitation, Q, the nature of Q , the frequency, f , and the 
system parameters m, c, and k. These parameters are 
customarily combined to describe the "natural frequency" 
as follows: 

f n = 27t v m (17-1) 



and the "damping ratio" as 

_ c 

D= 2Vkm 



(17-2) 



c. l Figure 17-B (a) shows the dynamic response 
of the system when the amplitude of the exciting force, 
Q , is constant. The abscissa of the diagram is the 
dimensionless ratio of exciting frequency, f , divided by 
the natural frequency, f n , in equation (17-1). The 
ordinate is the dynamic magnification factor, M z , which is 
the ratio of A z to the static displacement, A z = (Qo/k). 
Different response curves correspond to different values 
of D. 

d. l Figure 17-E (b) is the dynamic response of 
the system when the exciting force is generated by a 
rotating mass, whichdevelops: 

Q = m e (e)47i 2 f 2 (17-3) 



17-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



where 



m e =the total rotating mass 
e = the eccentricity 

f = the frequency of oscillation, cycles per 
second 



e. The ordinate M z . I (fig YPk ib)) relates the 
dynamic displacement, A z , to m e e/m. The peak value of 
the response curve is a function of the damping ratio and 
is given by the following expression: 

1 



M 



z(max) 



orM z = 2DVl-D" 



(17-4) 



For small values of D, this expression becomes 1/2D. 
These peak values occur at frequency ratios of 

fo 

f n =Vl-D 2 Iff ig. 17-26 ) 
or (17-5) 

fo J 

f n = V1-2D 2 Kfia. 17-2b ) 



17-3. Foundations on elastic soils. 

a. Foundations on elastic half-space. For very 
small deformations, assume soils to be elast ic materials 
with properties as noted in Iparaaraph 3-8. 1 Therefore, 
theories describing the behavior of rigid foundations 
resting on the surface of a semi-infinite, homogeneous, 
isotropic elastic body have been found useful for study of 
the response of real footings on soils. The theoretical 
treatment involves a circular foundation of radius, r , on 
the surface of the ideal half-space. This foundation has 
six degrees of freedom: (1-3) translation in the vertical 
(z) or in either of two horizontal (x and y) directions; (4) 
torsional (yawing) rotation about the vertical (z) axis; or 
(5-6) rocking (pitching) rotation about either of the two 
horizontal ( x and y) axe s. These vibratory motions are 
illustrated in lfiqure 17^31 

(1) A significant parameter in evaluating 
the dynamic response in each type of motion is the 
inertia reaction of the foundation. For translation, this is 
simply the mass, m = (W/g); whereas in the rotational 




V 



) 



).o 10 

FREQUENCY, CPS 



(Courtesy) of F. E Richart, Jr., J R. Hall. Jr., and R. D. 
Woods, Vibrations of Soils and Foundations , 1970, p 316. 
Reprinted by permission of Prentice-Hall, Inc., Englewood 
Cliffs, N. J.) 



Figure 17-1. Response spectra for tibraton limits. 
17-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



90 
40 
90 

X0 



— i r- 

Const. 



< 9 

i 

Z 




0.1 



O Ol2 0.4 O* 



10 U 1.4 I* l.t UO 



(a) CONST. FORCE AMPLITUDE EXCITATION 
Q « o sin u«t 




02 04 OA OJ 1.0 I.Z 1.4 l.t t* ii> 

(b) ROTATING MASS EXCITATION 
Q * in ew sinuJt 



(Courtesy of F. E. Richart, Jr., J. R. Hall, Jr., and R. D. Woods, 
Vibrations of Soils and Foundations , 1970, pp 383-384. Reprinted by 
permission of Prentice-Hall, Inc., Englewood Cliffs, N. J.) 

Figure 1 7-2. Response curves for the single-degree-of-freedom system with viscous damping. 




U. S. Army Corps of Engineers 



Figure 1 7-3. Six modes of vibration for a foundation. 



Figure 17-3. Six modes of vibration for a foundation. 



17-3 



TM 5-818-1 / AFM 88-3, Chap. 7 



modes of vibration, it is represented by the mass 
moment of inertia about the axis of rotation. For torsional 
oscillation about the vertical axis, it is designated as I e ; 
whereas for rocking oscillation, it is I ¥ , (for rotation about 
the axis through a diameter of the base of the 
foundation). If the foundation is considered to be a right 
circular cylinder of radius r , height h, and unit weight y, 
expressions for the mass and mass moments of inertia 
are as follow: 



n r hy 



m = g 

n r 4 h y 
ls= 2g 

OTn 2 hv ( r 2 + 

\= g 4 



*-) 



(17-6) 

(17-7) 
(17-8) 



(2) Theoretical solutions describe the 
motion magnification factors M, or ML, for example, in 
terms of a "mass ratio" B z a nd a dimensionless 
frequency factor a I Table 1 TA\ lists the mass ratios, 
damping ratios, and spring constants corresponding to 
vibrations of 



the rigid circular footing resting on the surface of an 
elastic semi-infinite body for each of the modes of 
vi bration. Introduce these quantities into equations given 
in lparagraph 17-2 t o compute resonant frequencies and 
amplitudes of dynamic motions. The dimensionless 
frequency, ao, for all modes of vibration is given as 
follows: 



27rf r 



a = V s 



/E 
= cor V G 



(17-9) 



The shear velo city, V s , in the soil is discussed in 

I paragraph 17-51 

(3) 1 Figure 17-4 shows the variation of the 
damping ratio, D, with the mass ratio, B, for the four 
modes of vibration. Note that D is significantly lower for 
the rocking mode than for the vertical or horizontal 
translational modes. Using the expression M = 1/(2D) 
for the amplitude magnification factor and the 
appropriate D, fro m 1 figure 17-41, it is obvious that M, can 
become large. For example, if B v = 3, then D v = 0.02 
andM = 1/(2x0.02) = 25. 



Table 17-1. Mass ratio, Damping Ration, and Spring Constant for Rigid Circular Footing on the Semi-Infinite Elastic Body 



Mode of 
Vibration 

Vertical 



Mass (or Inertia) 

Ratio, B. 
' l 



(1 - v) 



P r " 



Damping Ratio 

D. 



D - 
z 



0.425 



Spring Con- 
stant k. 
i 

4Gr 

o_ 

*z ~ 1 - V 



Sliding 



B = 
x 



(7 - 8v)m 
32(1 - v)pr" 



0.288 



X J£ 



= 32(1 - V) 
x 7 - 8V o 



Rocking 



_ 3(1 - v) _jt 

\ ~ 8 5 

pr 
r o 



0.15 



(1 



+ V^ 



k, = 



8 Gr- 



ill 3(1 - v) 



Torsional B„ = 



pr 



0.50 



1 + 2B 



6 



k e = 



^Gr 3 
3 o 



U. S. Army Corps of Engineers 



17-4 



TM 5-818-1 / AFM 88-3, Chap. 7 




«,«• 



•»•.,%,«■, 



(Courtesy of F. E. Richart, Jr. J. ft Hall, Jr., and R. D. 
Woods, Vibrations of Soils and Foundations , 1970. p 
226. Reprinted by permission of Prentice-Hall, Inc., 
Englewood Cliffs, N. J.) 

Figure 1 7-4. Equivalent damping ratio for oscillation of 
rigid circular footing on elastic half-space. 

b. Effects of shape of foundation. The 
theoretical solutions described above treated a rigid 
foundation with a circular contact surface bearing against 
the elastic half-space. However, foundations are usually 
rectangular in plan. Rectangular footings may be 
converted into an equivalent circular footing having a 
radius ro determined by the following expressions: 

For translation in z- or x-directions: 

r D ^ /ScSZ (17-10) 



For rocking: 



r =4 A 6cd" 



3n 

For torsion: 

r n =4 /T6cd(c" + & 



(17-11) 



(17-12) 



671 



In equations (17-10), (17-11), and (17-12), 2c is the 
width of the rectangular foundation (along the axis of 
rotation for rocking), and 2d is the length of the 
foundation (in the plane of rotation for rocking). Two 
values of r are obtained for rocking about both x and y 
axes. 

c. Computations. I Figure 17-5 1 presents 
examples of computations for vertical motions (Example 
1) and rocking motions (Example 2). 

d. Effect of embedment. Embedment of 
foundations a distance d below the soil surface may 
modify the dynamic response, depending upon the soil- 
foundation contact and the magnitude of d. If the soil 
shrinks away from the vertical faces of the embedded 
foundation, no beneficial effects of embedment may 
occur. If the basic evaluation of foundation response is 
based on a rigid circular footing (of radius r ) at the 
surface, the effects of embedment will cause an increase 



in resonant frequency and a decrease in amplitude of 
motion. These changes are a function of the type of 
motion and the embedment ratio d/r . 

(1) For vertical vibrations, both analytical 
and experimental results indicate an increase in the 
static spring constant with an increase in embedment 
depth. Embedment of the circular footing a distance d/r 
< 1 .0 produces an increase in th e embedd ed spring 
constant k zd < which is greater than k z rtable 17-1 ) by k zd /k z 
s (1 + 0.6 d/r ). An increase in damping also occurs, i.e., 
D zd /D z (1 + 0.6 d/r ). These two approximate relations 
lead to an estimate of the reduction in amplitude of 
motion because of embedment from A zd / A z = 1 / D zd / D z 
x k zd /k z ). This amount of amplitude reduction requires 
complete soil adhesion at the vertical face, and test data 
have often indicated less effect of embedment. Test 
data indicate that the resonant frequency may be 
increased by a factor up to (1 + 0.25 d/r ) because of 
embedment. 

(2) The influence of embedment on 
coupled rocking and sliding vibrations depends on the 
ratio B m /B x (l iable 17-1)1 For B B /B X = 3.0, the increase in 
natural frequency due to embedment may be as much as 
(1 + 0.5 d/r ). The decrease in amplitude is stongly 
dependent upon the soil contact along the vertical face of 
the foundation, and each case should be evaluated on 
the basis of local soil and construction conditions. 

e. Effect of finite thickness of elastic layer. 
Deposits of real soils are seldom homogeneous to 
significant depths; thus theoretical results based on the 
response of a semi-infinite elastic media must be used 
with caution. When soil layers are relatively thin, with 
respect to foundation dimensions, modifications to the 
theoretical half-space analyses must be included. 

(1) Generally, the effect of a rigid layer 
underlying a single elastic layer of thickness, H, is to 
reduce the effective damping for a foundation vibrating at 
the upper surface of the elastic layer. This condition 
results from the reflection of wave energy from the rigid 
base back to the foundation and to the elastic medium 
surrounding the foundation. For vertical or torsional 
vibrations or a rigid circular foundation resting on the 
surface of the elastic layer, it has been established that a 
very large amplitude of resonant vibrations can occur if 



V s 



> 41-1(17-13) 



fo 



In equation (17-13), V, is the shear wave velocity in the 
elastic layer and fo is the frequency of footing vibrations. 
When the conditions of equation (17-4) occur, the natural 
frequency (equation (17-1)) becomes the important 
design criterion because at that frequen- 



17-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



EXAMPLE 1 
a. foundation for single - cylinder 
vertical compressor 



EXAMPLE 2 

B. MACHINE FOUHDATXOH SUBJECTED TO 
ROCKING VIBRATXONMJ 




14" BOM , 9 

430 mm opbmtim speed 

unbalanced forces i 

vertical primary - 7720 lb 
vertical secondary - 1885 lb 
rom*. primary - 104 lb 
boris. secondary - lb 



NT. MACHINE AND 
MOTOR - 10900 LB. 



design criterion i smooth operatic* 

( LESS TRAM 0.10 XH / BBC VELOCITY ) 
AT 450 KM THIS REQUIRES A_ m 0.002 XH. 



SOIL PROPERTIES! ▼, - 680 FT / SBC 

G - 11,000 LB / I* 2 
4* m 110 LB / FT 3 
V - 0.33 

SOLUTION i 

FOR FIRST ESTIMATE OF FOUNDATION! SIZE, 
DETERMINE STATIC SIZE FOR 



a m - 01002 IH. 



'«■ 



72.8" 



0.002* - 



(i-v'Q o.*«7 imo+mt) 



4 O r„ 



y, /<<*»*/; 



6.07 FT FOR CIRCULAR FOUNDATION 



THEN REOUIRED AREA -NT 2 - 115.6 FT 3 

TRY 15'x8'ic3' THICK FOUNDATION BLOCK 
»nd r ■ 6. IB FT. 



THEN A - 120 FT' 



WT. FOUNDATION BLOCK - 54,000 LB. 
NT. TOTAL -Mr 64,900 LB. 

FROM TABLE 17-1 i 

tl-V) * 0.67 X 64900 . .. 
B « " 4?z£ * 4 x 110 (6.18^ " ° 42 



0.425 



M ( » l.o 
(1-y) O p 
4 G r_ 



0.66 



THUS A.- A, 



■« 0.00197' 



■a 
FOR 



6.18 FT. 



THEREFORE, THE 15'x 8'x 3' THICK CONCRETE 

BLOCK FOUNDATION IS SATISFACTORY 



U. S. Army Corps of Engineers 



18' 













tWSSOOOO LB. 

I .2.88 x 10* 
rr.LB.88C* 



^x 



DESIGN CRITERIOH i 0.20 IN / SBC BOM 
MOTION AT MACHINE 



SOIL 



AT 1300 RPM THIS LIMITS A^ TO 0.0013 XH 

FROM COMBINED ROCKXNO AHO SLXDXMG. 

(AT SLOWER SPEEDS THE ALLOWABLE A_ 
IS LAROBR ) 

PROPERTIES i T B - 770 FT / BBC 



S - 14,000 LB / XH 2 
/* - 110 LB / FT S 
V - 0.33 
HORIZONTAL TRANSLATION OHLYi 



EQUIVALENT 









Oe (7 - 8y) 
32 (1-V) O r. 



1.0 



• 0.00003 XH. 



HORIZONTAL TRAHSLATIOH IS HEOLXOXBLB 
ROCKIMQ ABOUT 
EOUIVALENT r„ -I P* - V - ™'"" - - 12.04 FT 



P JfWW ° , 

.Tfijcdf _ J 734(18) 3 _ 
V 3ir * 3» 



3 (l-v) I 



THEN D- 



0.15 



3(0.67)2.88x10* 

8 



• 0.8] 



(l+V -/By 



10(12.04)9 
. 0.09 , AND FROM 



EQ.17-4, M w - 5.6 



THE STATIC MOMENT ABOUT O IS 

T B - 400 X 18 - 720O FT. LB. , AMD THE 
STATIC ANBULAR DEFLECTION IS 

7200 X 3(0.67) 0.51 



V* "^ " 8(14000) 144(12.04)* " 10* 



RAD 



THIS ROTATION MOULD PRODUCE A HORIZONTAL 
MOTION AT THE MACHINE CENTERLINE OF 

JL.i.}- (18 X 12) - 1. 10 X lO^IN. 
10° 
OR, THE DYNAMIC AMPLITUDE AT RESONANCE IS 

■ 6.17 X 10~* IN. < 0.0015 IN. 



*x. -fi h 

DYNAM 
A,, -M.. A 



« 



Figure 1 7-5. Examples of computations for vertical and rocking motions. 



17-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



cy excessive dynamic motion will occur. To restrict the 
dynamic oscillation to slightly larger than the static 
displacement, the operating frequency should be 
maintaine d at one half, or less, of the natural frequency 
If fig 17-21 . 

(2) The relative thickness (expressed by 
H/ro) also exerts an important influence on foundation 
response. If H/r is greater than about 8, the foundation 
on the elastic layer will have a dynamic response 
comparable to that for a foundation on the elastic half- 
space. For H/r < 8, geometrical damping is reduced, 
and the effective spring const ant is incre ased. The 
values of spring cons tant, k, in Itable 17-1 l are taken as 
reference values, and l table 1 7-2 1 indicates the increase 
in spring constant associated with a decrease in 
thickness of the elastic layer. Values of the increase in 
spring constants for sliding and for rocking modes of 
vibration will tend to fall between those given for vertical 
and torsion for comparable H/r conditions. 

f. Coupled modes of vibration. In general, 
vertical and torsional vibrations can occur independently 
without causing rocking or sliding motions of the 
foundation. To accomplish these uncoupled vibrations, 
the line of action of the vertical force must pass through 
the center of gravity of the mass and the resultant soil 
reaction, and the exciting torque and soil reaction torque 
must be symmetrical about the vertical axis of rotation. 
Also, the center of gravity of the foundation must lie on 
the vertical axis of torsion. 

(1) When horizontal or overturning 
moments act on a block foundation, both horizontal 
(sliding) and rocking vibrations occur. The coupling 
between these motions depends on the height of the 
center of gravity of. the machine-foundation about the 
resultant soil reaction. Details of a coupl ed rocking a nd 
sliding analysis are given in the example i nl figure 17^61 

(2) A "lower bound" estimate of the first 
mode of coupled rocking and sliding vibrations can be 
obtained from the following: 



1 = 1 + 1 

,2 f 2 ,2 
fo fx +f„ 



(17-14) 



In equation (17-14), the resonant frequencies in the 
sliding x and rocking v|/ motions can be determined by 



introducing values from Itable 17-1 1 into equations (17-1) 
and (17-5). (Note that equation (17-14) becomes less 
useful when D x is greater than about 0.15). The first 
mode resonant frequency is usually most important from 
a design standpoint. 

g. Examples. I Figure 17-5,1 Example 1 , 
illustrates a procedure for design of a found ation to 
suppo rt machine-producing vertical excitations. I Figurel 
1 1 7-5] Example 2, describes the analysis of uncoupled 
horizontal and rocking motion for a particular foundation 
subjected to horizontal excitations. The design 
procedure of Example 1 is essentially an iterative 
analysis after approximate dimensions of the foundation 
have been established to restrict the static deflection to a 
value comparable t o the design c riterion. 

(1) In l figure 17-5,1 Example 1 shows that 
relatively high values of damping ratio D are developed 
for the vertical motion of the foundation, and Example 2 
illustrates that the high damping restricts dynamic mo- 
tions to values slightly larger than static displacement 
caused by the same force. For Example 2, establishing 
the static displacement at about the design limit value 
leads to satisfactory geometry of the foundat ion. 

(2) Example 2 Iffig 1 7-5)1 gives the 
foundation geometry, as well as the analysis needed to 
ascertain whether the design criterion is met. It is 
assumed that the 400-pound horizontal force is constant 
at all frequencies and that a simple superposition of the 
singledegree-of-freedom solutions for horizontal 
translation and rocking will be satisfactory. Because the 
horizontal displacement is negligible, the rocking motion 
dominates, with the angular rotation at resonance 
amounting to (M v x v|/ s ) or A ¥ = 5.6 x 0.51 x 10" 6 = 2.85 x 
10" 6 radians. By converting this motion to horizontal 
displacement at the machine center line, it is found that 
the design condition s are met. 

(3) In ffigure 17^6] the foundation of 
Example 2 (fig. 17-51 is analyzed as a coupled system 
including both rocking and sliding. The response curve 
for angular rotation shows a peak motion of A^ = 2.67 x 
10" 6 radians, which is comparable to the value found by 
considering rocking alone. The coupled dynamic 
response of any rigid foundation, e.g., a radar tower, can 



Table 1 7-2. Values of k L /L for Elastic Layer (k from TabJeJZJ} 



H/r 

Vertical 

Torsion 



0.5 
5.0 



1.0 
2.2 
1.07 



2.0 

1.47 

1.02 



4.0 

1.23 

1.009 



8.0 
1.10 



1.0 
1.0 



U. S. Army Corps of Engineers 



17-7 



IN THE SKETCH REPRESENTING THE DYNAMIC MOTION OF 
THE FOUNDATION OF FIGURE 17-5, EXAMPLE 2, THE 
SUBSCRIPT "g" REFERS TO THE CENTER OF GRAVITY, 
AND "b" REFERS TO THE CENTER OF THE BASE. 

2 



x b " "9 ~ h ot* x b 

a - TOTAL MASS, AND I 



Ik - I g ♦ 



°* 2 VsT»" 



MASS MOMENT OP INERTIA 

FROM EQUATION 17-2 



18.4 (l-v) 2 nrz 
(7 - evj'o »^ G 



c„ * d- 2 VICT - ?: B0 ,T°, *£\ 



P„ - - c_ 



x b - *x 



- c v . 



AND VALUES FOR B 
O., and k 1 
TABLE 17-1 



i' 



TM 5-818-1 / AFM 88-3, Chap. 7 



4* 



A 






^4U 



V 

LL 



Lit 



/««WiwW^JWJJw. 



(M 



(AM 



L 






THE EQUATION OF EQUILIBRIUM FOR HORIZONTAL TRANSLATION IS 



■ *, + C X 



*X X b 



- 0- 



■ x b ♦ m h y ♦ c„ x b + kx x b 



AND FOR ROTATION ABOUT THE CENTER OF GRAVITY IT IS 



I gT v + «Vf + V? 



*b ~ *x h o x b " T » 



OR 



I K y? - m h 2 ^ + c„(f> ♦ k^jK - c x x b h„ - V, x b h - T„ 



© 



© 



© 



LET Xv » &X1 ainu/t ♦ Ajjj count ■ A x ain {utt -«„) 
y» - A^ alnwt + A n coau»t - Ap, ain («vt -«V> 
O - Q,, ainut j NOTEj f^m ^A^ + A^j ; tanST* . . 

*r" V**i + *yj i tan** - -a w Arj 



INTRODUCING THE EXPRESSIONS @ INTO EQUATIONS. @ AND (b) GIVE FOUR EQUATIONS 
WITH FOUR UNKNOWNS (A„l , A x 2 , Af 1 , A* 2 ' ' t13R EACH CHOSEN VALUE OF u 
(u - 277-FREQUENCY ) . THUS A COMPUTER SOLUTION IS NEEDED. THE GRAPH BELCH 
SHOWS THE ROCKING RESPONSE CURVE FOR THE FOUNDATION (SEE SKETCH ABOVE AND 
FIGURE 17-5) . THE PARAMETERS NEEDED FOR THE SOLUTION ARE NOTED BELOW. 



o - 400 LB. 

h - 18 FT. 

550,000 
" _ 32.2 



(FREQUENCY INDEPENDENT) 1 
h Q - XI FT. 

. 17,080 LB SEC 2 / FT. 



I b - 2.88 X 10 6 FT LB SEC 2 

r - 13.96 FT ( SLIDING ) 

r Q - 12.04 FT ( ROCKING > 

J, ■ 1,39 « 10 8 LB / FT 

k* - 1.41 X 10 10 FT LB / RAD 

C x - 1.45 X 10 6 LB SEC / FT 

Cy - 3.62 X 10 7 FT LB SVG//tA0 

FOR ROTATING MASS MACHINE 
TYPE EXCITATION, WE WOULD 
INTRODUCE 



J. 



°o " "• • u)1 

eccentric mass 
eccentric radius 




t cmiiitc I 



U. S. Army Corps of Engineers 

Figure 1 7-6. Coupled rocking and sliding motion of foundation. 



17-8 



TM 5-818-1 / AFM 88-3, Chap. 7 



be evaluated by the procedure illustrated ii j figure 17^61 

17-4. Wave transmission, attenuation, and 

isolation. Vibrations are transmitted through soils by 
stress waves. For most engineering analyses, the soil 
may be treated as an ideal homogeneous, isotropic 
elastic material to determine the characteristics of the 
stress waves. 

a. Half-space. Two types of body waves may 
be transmitted in an ideal half-space, compression (P-) 
waves and shear (S-) waves; at the surface of the 
halfspace, a third wave known as the Rayleigh (R-) wave 
or surface wave will be transmitted. The characteristics 
that distinguish these three waves are velocity, wavefront 
geometry, r adiation damping, and particle motion. 
I Figure 17T7"| shows the characteristics of these waves as 
they are generated by a circular footing undergoing 
vertical vibration on the surface of an ideal half-space 
with is |x = 0.25. The distance from the footing to each 
wave in l figure 17-71 is drawn in proportion to the velocity 
of each wave. The wave velocities can be computed 
from the following: 

P 

P-wave velocity: 
v c = /Xh2G (17-15) 

V P 
S-wave velocity: 



v s =/ G 



v-p- 

R-wave velocity: 
v R = Kv s 



where 



(17-16) 



(17-17) 



A. = 2uG and G are Lame's E 

1-2|j. constants; G =2(1 + j) 

p = y/G = mass density of soil 

y = moist or saturated unit weight 

K = constant, depending on Poisson's ratio 
0.87 ^K ^0.98 for < < 0.5 

(1) The P- and S-waves propagate radially 
outward from the source along hemispherical wave 
fronts, while the R-wave propagates outward along a 
cylindrical wave front. All waves encounter an 
increasingly larger volume of material as they travel 
outward, thus decreasing in energy density with 
distance. This decrease in energy density and its 
accompanying decrease in displacement amplitude is 
called geometrical clamping or radiation damping. 

(2) The particle motions are as follows: for 
the P-wave, a push-pull motion in the radial direction; for 
the S-wave, a transverse motion normal to the radial 
direction; and for the R-wave, a complex motion, which 
varies with depth and which occurs in a vertical plane 
containing a radius. At the surface, R-wave particle 
motion describes a retrogra de ellipse. The shaded 
zones along the wave fronts in | figure 17-7| represent the 



relative particle amplitude as a function of inclination 
from vertical. 

b. Layered media. 

(1) In a layered medium, the energy 
transmitted by a body wave splits into four waves at the 
interface between layers. Two waves are reflected back 
into the first medium, and two waves are transmitted or 
refracted into the second medium. The amplitudes and 
directions of all waves can be evaluated if the properties 
of both media and the incident angle are known. If a 
layer containing a lower modulus overlies a layer with a 
higher modulus within the half-space, another surface 
wave, known as a Love wave, will occur. This wave is a 
horizontally oriented S-wave whose velocity is between 
the S-wave velocity of the layer and of the underlying 
medium. 

(2) The decay or attenuation of stress 
waves occurs for two reasons: geometric or radiation 
damping, and material or hysteretic damping. An 
equation including both types of damping is the following: 

A 2 =A! t+ C exp[-a (r 2 - n )] (17-18) 

r 2 
where 

A 2 = desired amplitude at distance r2 
A, = known or measured amplitude at 

radial distance r, from vibration 

source 
C = constant, whichdescribes 

geometrical damping 

1 for body (P- or S-) waves 

0.5 for surface or R-waves 
a = coefficient of attenuation, which 

d escribes ma terial damping (values 

in ltable 1 7-3| 

c. Isolation. The isolation of certain structures 
or zones from the effects of vibration may sometimes be 
necessary. In some instances, isolation can be 
accomplished by locating the site at a large distance 
from the vibration source. The required distance, r 2 , is 
calculated.from equation (17-18). In other situations, 
isolation may be accomplished by wave barriers. The 
most effective barriers are open or void zones like 
trenches or rows of cylindrical holes. Somewhat less 
effective barriers are solid or fluid-filled trenches or 
holes. An effective barrier must be proportioned so that 
its depth is at least two-thirds the wavelength of the 
incoming wave. The thickness of the barrier in the 
direction of wave travel can be as thin as practical for 
construction considerations. The length of the barrier 
perpendicular to the direction of wave travel will depend 
upon the size of the zone to be isolated but should be no 
shorter than two times the maximum plan dimension of 
the structure or one wavelength, whichever is greater. 
17-5. Evaluation of S-wave velocity In soils. The 
key parameter in a dynamic analysis of a 



17-9 



TM 5-818-1 / AFM 88-3, Chap. 7 



Circular Footing 



r -2 Geometrical r -2 ( -05 
^ Damping Law -*_ ^ 




Relative 
Amplitude 



Shear 
Window 



(Courtesy of F. E. Richart, Jr., J. R. 
Vibrations of Soils and Foundations, 



Hall, Jr., and R. D. Woods. 
1970, p 91. Reprinted by 



permission of Prentice-Hall, Inc., Englewood Cliffs, N. J.) 
Figure 17-7. Distribution of displacement waves from a circular footing on the elastic half-space. 



soil-foundation system is the shear modulus, G. The 
shear modulus can be determined in the laboratory or 
estimated by empirical equations. The value of G can 
also be computed by the field-measured S-wave velocity 
and equation (17-16). 

a. Modulus at low strain levels. The shear 
modulus and damping for machine vibration problems 
correspond to low shear-strain amplitudes of the order of 
1 to 3 x 10" 4 percent. These properties may be 
determined from field measurements of the seismic 



wave velocity through soil or from special cyclic 
laboratory tests. 

b. Field wave velocity tests. S-wave velocity 
tests are preferably made in the field. Measurements 
are obtained by inducing a low-level seismic excitation at 
one location and measuring directly the time required for 
the induced S-wave to travel between the excitation and 
pickup unit. Common tests, such as uphole, downhole, 
or crosshole propagation, are described in geotechnical 
engineering literature. 



Table 17-3. Attenuation Coefficients for Earth Materials 



Materials 



Sand Loose, fine 
Dense, fine 



a(1/ft)@50HZ a 

0.06 
0.02 



Clay Silty (loess) 
Dense, dry 

Rock Weathered volcanic 
Competent marble 



0.06 
0.003 

0.02 
0.00004 



a a is a function of frequency. For other frequencies, f, compute cc f = (f/50) x a 50 



U. S. Army Corps of Engineers 



17-10 



TM 5-818-1 / AFM 88-3, Chap. 7 



(1) A problem in using seismic methods to 
obtain elastic properties is that any induced elastic pulse 
(blast, impact, etc.) develops three wave types previously 
discussed, i.e., P-, S-, and R-waves. Because the 
velocity of all seismic waves is hundreds of feet per 
second and the pickup unit detects all three wave pulses 
plus any random noise, considerable expertise is 
required to differentiate between the time of arrival of the 
wave of interest and the other waves. The R-wave is 
usually easier to identify (being slower, it arrives last; 
traveling near the surface, it contains more relative 
energy). Because R- and S-wave velocities are relatively 
close, the velocity of the R-wave is frequently used in 
computations for elastic properties. 

(2) Because amplitudes in seismic survey 
are very small, the computed shear and Young's moduli 
are considerably larger than those obtained from 
conventional laboratory compression tests. 

(3) The shear modulus, G, may be 
calculated from the S- (approximately the R-wave) wave 
velocity as follows: 



G = pV s 



(17-19) 



where 
P = 



y/32.2 = mass density of soil using 

wet or total unit weight 
V s = S-wave velocity (or R-wave), feet per 

second 
This equation is independent of Poisson's ratio. The V s 
value is taken as representative to a depth of 
approximately one-half wavelength. Alternatively, the 
shear modulus can be computed from the P-wave 
velocity and Poisson's ratio from: 

G = P( 1 - 2u)V c 2 (17-20) 

2(1- n) 
The use of this equation is somewhat limited because 
the velocity of a P-wave in water is approximately 5000 
feet per second (approximately the velocity in many 
soils) and Poisson's ratio must be estimated. For 
saturated or near saturated soils, n - 0.5. The theoretical 
variation of the ratio V s /V p with p. is shown in figure 17-81 



> 

d 

F 3 
< 

a. 

U 

o 

-J 2 

111 

> 

> 
< 

s 



v 


1 


- 




p 

V 

s 




V = 

s 


V7 







































O O.I 0.2 0,3 0.4 0.5 

POISSONCS RATIO 

U. S. Army Corps of Engineers 

Figure 1 7-8. Theoretical relation between shear velocity ratio Vp/Vs and Poisson's ratio. 

17-11 



TM 5-818-1 / AFM 88-3, Chap. 7 



c. Laboratory measurement of dynamic 
stress-strain properties. Low shear-strain amplitude, i.e. 
less than 10" 2 percent, shear modulus data may be 
obtained from laboratory tests and usually involve 
applying some type of high-frequency forced vibration to 
a cylindrical sample of soil and measuring an appropriate 
response. Some types of tests allow the intensity level of 
the forced vibration to be varied, thus yielding moduli at 
different shear strains. 

(1) High strain-level excitation, i.e. 0.01 to 
1.0 percent, may be achieved by low-frequency, cyclic 
loading triaxial compression tests on soil samples. The 
modulus, damping, and strain level for a particular test 
are calculated directly from the sample response data. 
The usual assumption for calculating the modulus and 
damping from forced cyclic loading tests on laboratory 
samples is that at any cyclic strain amplitude the soil 
behaves as a linear elastic, viscous, damped material. 
A typical set of re sults may take the form of a hysteresis 
loop as shown ir l figure 17-91 Either shear or normal 
stress cyclic excitation may be used. The shear modulus 
is calculated from the slope of the peak-to-peak secant 
line. The damping is computed from the area of the 



hysteresis loop, and the strain level is taken as the 
single-amplitude (one-half the peak-to-peak amplitude or 
origin to peak value) cyclic strain for the condition during 
that cycle of the test. Not e that the equa tions for modulus 
and damping shown in I figure 17-9 l assume the soil 
behaves as an equivalent elastic viscous, dampened 
material, which is linear within the range of strain 
amplitude specified. This assumption is usually made in 
most soil dynamics analyses because of the low- 
vibration amplitudes involved. If the cyclic hysteresis 
loops are obtained from triaxial test specimens, the 
resulting modulus will be the stress-strain modulus, E. If 
the tests involve simple shear or torsion shear such that 
shear stresses and strains are measured, the resulting 
modulus will be the shear modulus, G. In either case, 
the same equations apply. 

(2) The shear modulus, G, can be 
computed from the stress strain modulus and Poisson's 
ratio as follows: 



G = 



(17-21) 



2(1+1*) 




STRAIN 
VISCOUS DAMPING (/3) 

! ( AREA OF LOOP ACA'C' 



AREA OF TRIANGLES (OAB + OA B 



SINGLE AMPLITUDE STRAIN, £ 



BB' 



FORMULAS VALID FOR CYCLIC SHEAR STRESS 
(SINGLE SHEAR OR TORSION SHEAR TEST) OR 
CYCLIC AXIAL STRESS (TRIAXIAL TEST) 



U. S. Army Corps of Engineers 



Figure 1 7-9. Idealized cyclic stress-strain loop for soil. 



17-12 



TM 5-818-1 / AFM 88-3, Chap. 7 



The shear strain amplitude, A E , may be computed from 
the axial strain amplitude, E ,and Poisson's ratio as 
follows: 

A E = E(1+^) (17-22) 

For the special case of saturated soils, Poisson's ratio is 
0.5, which leads to the following: 
G = E/3 
A E =1.5 E 
d. Correlations. 

(1) Empirical correlations from many sets 
of data have provided several approximate methods for 
estimating the S-wave velocity and shear modulus for 
soils corresponding to low-strain excitation. For many 
undisturbed cohesive soils and sands: 

G =1230(21973 - e) 2 (OCR)" (o)0.5 (pounds 1 + per 
square inch) (17-23) 

where 

e = void ratio 

r\ = empirical constant, which depends on 
the PI of cohesive soils (|table 17-4U For sands, PI = 
and r\ = 0, so OCR term reduces to 1 .0. For clays, the 
maximum value is r| = 0.5 for PI > 1 00. 

a = 1/3 (ai + o 2 + 03) = mean normal 
effective stress, pounds per square inch 

(2) For sands and gravels, calculate the 
low-strain shear modulus as follows: 

G = 1OOO(K 2 )(0 O ) ' 5 (pounds per square foot) (17-24) 
where 

K 2 =empirical constant <table 17-51 

=90 to 190 for dense sand, gravel, and cobbles 
with little clay 

o = mean normal effective stress as in equation 
(17-23) (but in units of pounds per square foot) 



(3) For cohesive soils as clays and peat, 
the shear modulus is related to S u as follows: 

G = K 2 s u (17-25) 

For clays, K 2 ranges from 1 500 to 3000. For peats, K 2 
ranges from 1 50 to 1 60 (limited data base). 

(4) In the laboratory, the shear modulus of 
soil increases with time even when all other variables are 
held constant. The rate of increase in the shear modulus 
is approximately linear as a function of the log of time 
after an initial period of about 1000 minutes. The change 
in shear modulus, AG, divided by the shear modulus at 
1000 minutes, G 100 o, is called the normalized secondary 
increase. The normalized secondary increases range 
from nearly zero percent per log cycle for coarse sands 
to more than 20 percent per log for sensitive clays. For 
good correlation between laboratory and field 
measurements of shear modulus, the age of the in situ 
deposit must be considered, and a secondary time 
correction applies to the laboratory data. 

e. Damping in low strain levels. Critical 
damping is defined as 

c c = 2Vkm (17-26) 

where k is the spring constant of vibrating mass and m 
represents mass undergoing vibration (W/g). Viscous 
damping of all soils at low strain-level excitation is 
generally less than about 0.01 percent of critical damping 
for most soils or: 

D = c/c, <0.05 (17-27) 

It is important to note that this equation refers only to 
material damping, and not to energy loss by radiation 
away from a vibrating foundation, which may also be 
conveniently expressed in terms of equivalent viscous 
damping. Radiation damping in machine vibration 
problems is a function of the geometry of the problem 
rather than of the physical properties of the soil. 



Table 1 7-4. Values of Constant r Used with Equation (1 7-23) to Estimate Cyclic Shear Modulus at Low Strains 



Plasticity Index 



20 

40 

60 

80 

>100 



~T< 


0.18 
0.30 
0.41 
0.48 
0.50 



(Courtesy of 0. Hardin and P. Drnevich. "Shear 
Modulus and Damping in Soils: Design Equations and 
Curves," Journal., Soil Mechanics and Foundations 
Division . Vol 98. No. SM7. 1972, pp 667-692. Reprinted 
by permission of American Society of Civil Engineers, 
New York.) 



17-13 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 1 7-5. Values of Constant K2 Used with Equation (1 7-24) to Estimate Cyclic Shear Modulus at Low Strains for Sands 



e 

0.4 

0.5 

0.6 

0.7 

0.8 

0.9 



K 2 
70 
60 
51 
45 
39 
33 



D r {%) 

90 

75 

60 

45 

40 

30 



(Courtesy of H. B. Seed and L M. Idriss, "Simplified Procedures for 
Evaluating Liquefaction Potential/' Journal, Soil Mechanics and 
Foundations Division Vol 97, No. SM9. 1971, pp 1249-1273. Reprinted by 
permission of American Society of Civil Engineers, New York.) 



f. Modulus and damping at high strain levels. 
The effect of increasing ly higher strain levels is to reduce 
the modulus Kfig 17-10] and increase the damping of the 



soil jfiq 17-11)1 Shear modulus and damping values at 
high strains are used mainly in computer programs for 
analyzing the seismic response of soil under earthquake 
loading conditions. The various empirical relations for 
modulus and damping pertain to sands and soft, 
normally consolidated clays at low-to-medium effective 
confining pressures, in the range of about 100 feet or 
overburden. Stiff overconsolidated clays and all soils at 
high effective confining pressure exhibit lower values of 
damping and higher values of modulus, especially at 
high strain levels. As a maximum, the modulus and 
damping values for stiff or strong soils at very high 
effective confining pressures correspond to values 
pertaining to crystalline or shale-type rock. 

17-6. Settlement and liquefaction. 

a. Settlement. Repeated shearing strains of 
cohesionless soils cause particle rearrangements. 
When the particles move into a more compact position, 
settlement occurs. The amount of settlement depends 
on the initial density of the soil, the thickness of the 
stratum, and the intensity and number of repetitions of 
the shearing strains. Generally, cohesionless soils with 
relative densities (D r ) greater than about 75 percent 
should not develop settlements. However, under 10 6 or 
10 7 repetitions of dynamic loading, even dense sands 
may develop settlements amounting to 1 to 2 percent of 
the layer thickness. To minimize settlements that might 
occur under sustained dynamic loadings, the soil 
beneath and around the foundation may be 
precompacted during the construction process by 
vibroflotation, multiple blasting, pile driving, or vibrating 
rollers acting at the surface. The idea is to subject the 
soil to a more severe dynamic loading condition during 



construction than it will sustain throughout the design 
operation. 

b. Liquefaction of sands. The shearing 
strength of saturated cohesionless soils depends upon 
the effective stress acting between particles. When 
external forces cause the pore volume of a cohesionless 
soil to reduce the amount V, pore water pressures are 
increased during the time required to drain a volume V of 
water from the soil element. Consequently, pore 
pressure increases depend upon the time rate of change 
in pore volume and the drainage conditions (permeability 
and available drainage paths). When conditions permit 
the pore pressure, u, to build up to a value equal to the 
total stress, o n , on the failure plane, the shear strength is 
reduced to near zero and the mixture of soil grains and 
water behaves as a liquid. This condition is true 
liquefaction, in which the soil has little or no shearing 
strength and will flow as a liquid. Liquefaction or flow 
failure of sands involves a substantial loss of shearing 
strength for a sufficient length of time that large de- 
formations of soil masses occur by flow as a heavy 
liquid. 

c. Liquefaction due to seismic activity. Soil 
deposits that have a history of serious liquefaction 
problems during earthquakes include alluvial sand, 
aeolian sands and silts, beach sands, reclaimed land, 
and hydraulic fills. During initial field investigations, 
observations that suggest possible liquefaction problems 
in seismic areas include low penetration resistance; 
artesian heads or excess pore pressures; persistent 
inability to retain granular soils in sampling tubes; and 
any clean, fine, uniform sand below the groundwater 
table. The liquefaction potential of such soils for 
structures in seismic areas should be addressed unless 
they meet one of the criteria in Itable 17-61 In the event 
that 



17-14 



TM 5-818-1 / AFM 88-3, Chap. 7 



O.B 



0.6 















N^ 







































10" 



10 - ' 10 -1 

SINGLE-AMPLITUDE SHEAR STRAIN Y, PERCENT 



10 



(Courtesy of H. B. Seed and L M. Idriss, "Simplified Procedures for 
Evaluating Liquefaction Potential," Journal, Soil Mechanics and 
Foundations Division Vol 97, No. SM9. 1971, pp 1249-1273. Reprinted 
by permission of American Society of Civil Engineers, New York.) 

Figure 17-10. Variation of shear modulus with cyclic strain amplitude; G max = G atE = 1 to 3 x 10~ 4 percent; scatter in data 

up to about ± 0. 1 on vertical scale. 



17-15 



TM 5-818-1 / AFM 88-3, Chap. 7 



H 
Z 
U 
O 

<r 

LU 

D. 



I- 
< 

o 

z 

< 

Q 



30 



25 



20 



15 



10 





























/ ¥ 






J^ 


f 



























10 



10 



-3 



10 



-2 



10' 



SINGLE-AMPLITUDE SHEAR STRAIN Y, PERCENT 



10 



(Courtesy of H. B Seed and 1. M Idrisv. "Simplified 
Procedure for Evaluating Soil Liquefaction Potential." 
Journal, Soul Mechanics and Foundations Division. Vol 
97, No. SM9. 1971, pp 1249-1273. Reprinted by 
permission of the American Society, of Civil Engineers. 
Newt York.) 

Figure 17-11. Variation of viscous damping with cyclic strain amplitude, data scatter up to about ±50 percent of average 

dumping values shown for any strain. 



17-16 



TM 5-818-1 / AFM 88-3, Chap. 7 

Table 1 7-6. Criteria for Excluding Need for Detailed Liquefaction Analyses 

1. CL,CH,SC, or GC soils. 

2. GW or GP soils or materials consisting of cobbles, boulders, uniform rock fill, which have free-draining 
boundaries that are large enough to preclude the development of excess pore pressures. 

3. SP, SW, or SM soils which have average relative density equal to or greater than 85 percent, provided that 
the minimum relative density is not less than 80 percent. 

4. ML or SM soils in which the dry density is equal to or greater than 95 percent of the modified Proctor (CE 55) 
density. 

5. Soils of pre-Holocene age, with natural overconsolidation ratio equal to or greater than 16 and with relative 
density greater than 70 percent. 

6. Soils located above the highest potential groundwater table. 

7. Sands in which the "N" value is greater than three times the depth in feet, or greater than 75; provided that 75 
percent of the values meet this criterion, that the minimum "N" value is not less than one times the depth in feet, that there 
are no consistent patterns of low values in definable zones or layers, and that the maximum particle size is not greater 
than 1 in. Large gravel particles may affect "N" values so that the results of the SPT are not reliable. 

8. Soils in which the shear wave velocity is equal to or greater than 2000 fps. Geophysical survey data and site 
geology should be reviewed in detail to verify that the possibility of included zones of lowvelocity is precluded. 

9. Soils that, in undrained cyclic triaxial tests, under isotropically consolidated, stress-controlled conditions, and 
with cyclic stress ratios equal to or greater than 0.45, reach 50 cycles or more with peak-to-peak cyclic strains not greater 
than 5 percent; provided that methods of specimen preparation and testing conform to specified guidelines. 

Note: The criteria given above do not include a provision for exclusion of soils on the basis of grain-size distribution, and 
in general, grain-size distribution alone cannot be used to conclude that soils will not liquefy. Under adverse conditions 
nonplastic soils with a very wide range of grain sizes may be subject to liquefaction. 

U. S. Army Corps of Engineers 

17-17 



TM 5-818-1 / AFM 88-3, Chap. 7 



none of the criteria is met and a more favorable site 
cannot be located, the material in question should be 
removed, remedial treatment applied as described in 
I chapter 1"B , or a detailed study and analysis should be 
conducted to determine if liquefaction will occur. 



Ground motions from earthquakes cause motions of 
foundations by introducing forces at the foundation-soil 
contact zone. Methods for estimating ground motions 
and their effects on the design of found ation eleme nts 
are discussed in TM 5-809-1 / AFM 88-3, Chapter 18 . 



17-7. Seismic effects on foundations. 



17-18 



TM 5-818-1 / AFM 88-3, Chap. 7 



CHAPTER 18 



FOUNDATIONS IN AREAS OF SIGNIFICANT FROST PENETRATION 



18-1. Introduction. 

a. Types of areas. For purposes of this 
manual, areas of significant frost penetration may be 
defined as those in which freezing temperatures occur in 
the ground to sufficient depth to be a significant factor in 
foundation design. Detailed requirements of engineering 
design in s uch areas are given in TM 5-818-2/AFM 88-6, 
I Chapter 41 and the Arctic and Subarctic Construction 
series, TM 5-852-1 through 9/ AFM 88-19, Chapters 1 
through 9, respectively. Areas of significant frost 
penetration may be subdivided as follows: 

( 1 ) Seasonal frost areas. 

(a) Significant ground freezing occurs 
in these areas during the winter season, but without 
development of permafrost. 1 In northern Texas, 
significant seasonal frost occurs about 1 year in 10. A 
little farther north it is experienced every year. Depth of 
seasonal freezing increases northward with decreasing 
mean annual and winter air temperatures until 
permafrost is encountered. With still further decrease of 
air temperatures, the depth of annual freezing and 
thawing becomes progressively thinner. 

(b) The layer extending through both 
seasonal frost and permafrost areas in which annual 
freeze-thaw cycles occur is called the annual frost zone. 
In permafrost areas, it is also called the active layer. It is 
usually not more than 10 feet thick, but it may exceed 20 
feet. Under conditions of natural cover in very cold 
permafrost areas, it may be as little as 1 foot thick. Its 
thickness may vary over a wide range even within a 
small area. Seasonal changes in soil properties in this 
layer are caused principally by the freezing and thawing 
of water contained in the soil. The water may be 
permanently in the annual frost zone or may be drawn 
into it during the freezing process and released during 
thawing. Seasonal changes are also produced by 
shrinkage and expansion caused by temperature 
changes. 

(2) Permafrost areas. 

(a) In these areas, perennially frozen 
ground is found below the annual frost zone. In North 
America, permafrost is found principally north of latitudes 
55 to 65 degrees, although patches of permafrost are 
found much farther south on mountains where the 



Specialized terms relating to fro zen ground 
areas are defined in TM 5- 818-2/AFM 88-6, IChapter 4J 
and TM 5-852-1 /AFM 88-19. IChapte7TI 



temperature conditions are sufficiently low, including 
some mountains in the contiguous 48 States. In areas of 
continuous permafrost, perennially frozen ground is 
absent only at a few widely scattered locations, as at the 
bottoms of rivers and lakes. In areas of discontinuous 
permafrost, permafrost is found intermittently in various 
degrees. There may be discontinuities in both horizontal 
and vertical extent. Sporadic permafrost is permafrost 
occurring in the form of scattered permafrost islands. In 
the coldest parts of the Arctic, the ground may be frozen 
as deep as 2000 feet. 

(b) The geographical boundaries 

between zones of continuous permafrost, discontinuous 

permafrost, and seasonal frost without permafrost are 

poorly de- fined but are represented approximately in 

I figure 18-il 

b. General nature of design problems. 
Generally, the design of foundations in areas of only 
seasonal frost follows the same procedure as where 
frost is in- significant or absent, except that precautions 
are taken to avoid winter damage from frost heave or 
thrust. In the spring, thaw and settlement of frost- 
heaved material in the annual frost zone may occur 
differentially, and a very wet, poorly drained ground 
condition with temporary but substantial loss of shear 
strength is typical. 

(1) In permafrost areas, the same annual 
frost zone phenomena occur, but the presence of the 
underlying permafrost introduces additional potentially 
complex problems. In permafrost areas, heat flow from 
buildings is a fundamental consideration, complicating 
the design of all but the simplest buildings. Any change 
from natural conditions that results in a warming of the 
ground beneath a structure can result in progressive 
lowering of the permafrost table over a period of years 
that is known as degradation. If the permafrost contains 
ice in excess of the natural void or fissure space of the 
material when unfrozen, progressive downward thaw 
may result in extreme settlements or overlying soil and 
structures. This condition can be very serious because 
such subsidence is almost invariably differential and 
hence very damaging to a structure. Degradation may 
occur not only from building heat but also from solar 
heating, as under pavements, from surface water and 
groundwater flow, and from underground utility lines. 
Proper insulation will prevent degradation in some 
situations, but where a con 



18-1 



TM 5-818-1 / AFM 88-3, Chap. 7 



tinuous source of heat is available, thaw will in most 
cases eventually occur. 

(2) The more intense the winter cooling of 
the frozen layer in the annual frost zone and the more 
rapid the rate of frost heave, the greater the intensity of 
uplift forces in piles and foundation walls. The lower the 
temperature of permafrost, the higher the bearing 
capacity and adfreeze strength that can be developed, 
the lower the creep deformation rate under footings and 
in tunnels and shafts, and the faster the freeze-back of 
slurried piles. Dynamic response characteristics of 
foundations are also a function of temperature. Both 



natural and manufactured construction materials 
experience significant linear and volumetric changes and 
may fracture with changes in temperature. Shrinkage 
cracking of flexible pavements is experienced in all cold 
regions. In arctic areas, patterned ground is widespread, 
with vertical ice wedges formed in the polygon 
boundaries. When underground pipes, power cables, or 
foundation elements cross shrinkage cracks, rupture 
may occur during winter contraction. During summer 
and fall, expansion of the warming ground may cause 
substantial horizontal forces if the cracks have become 
filled with soil or ice. 




U. S. Army Corps of Engineers 



Figure 18-1. Frost and permafrost in North America. 
18-2 



TM 5-818-1 / AFM 88-3, Chap. 7 



(3) Engineering problems may also arise 
from such factors as the difficulty of excavating and 
handling ground when it is frozen; soft and wet ground 
conditions during thaw periods; surface and subsurface 
drainage problems; special behavior and handling 
requirements for natural and manufactured materials at 
low temperatures and under freeze-thaw action; possible 
ice uplift and thrust action on foundations; condensation 
on cold floors; adverse conditions of weather, cost, and 
sometimes accessibility; in the more remote locations, 
limited local availability of materials, support facilities, 
and labor; and reduced labor efficiency at low 
temperatures. 

(4) Progressive freezing and frost heave 
of foundations may also develop under refrigerated 
ware- houses and other facilities where sustained interior 
below-freezing temperatures are maintained. The 
design procedures and technical guidance outlined in 
this chapter may be adapted to the solution of these 
design problems. 

18-2. Factors affecting design of foundations. 

a. Physiography and geology. Physiographic 
and geologic details in the area of the proposed 
construction are a major factor determining the degree of 
difficulty that may be encountered in achieving a stable 
foundation. For example, pervious layers in fine-grained 
alluvial deposits in combination with copious 
groundwater supplies from adjacent higher terrain may 
produce very high frost-heave potential, but clean, free- 
draining sand and gravel terrace formations of great 
depth, free of excess ice, can provide virtually trouble- 
free foundation conditions. 

b. Temperature. The most important factors 
contributing to the existence of adverse foundation 
conditions in seasonal frost and permafrost regions are 
cold air temperatures and the continual changes of 
temperature between summer and winter. Mean annual 
air temperatures usually have to be 2- to 8 S F below 
freezing for permafrost to be present, although 
exceptions may be encountered both above and below 
this range. Ground temperatures, depths of freeze and 
thaw, and thickness of permafrost are the product of 
many variables including weather, radiation, surface 
conditions, exposure, snow and vegetative cover, and 
insulating or other special courses. The properties of 
earth materials that determine the depths to which 
freezing-and-thawing temperatures will penetrate below 
the ground surface under given temperature differentials 
over a given time are the thermal conductivity, the 
volumetric specific heat capacity, and the volumetric 
latent heat of fusion. These factors in turn vary with the 
type of material, density, and moisture content. I Figurel 

1 18-2 1 shows how ground temperatures vary during the 
freezing season in an area of substantial seasonal 
freezing having a mea n annual te mperature of 37 a F 
(Limestone, Maine), an d figure 18-3l shows similar data 



for a permafrost area having a mean annual temperature 
of 26 a F (Fairbanks, Alaska). 

(1 ) For the computation of seasonal freeze 
or thaw penetration, freezing-and-thawing indexes are 
used based upon degree-days relative to 32 F. For the 
average permanent structure, the design indexes should 
be those for the coldest winter and the warmest summer 
in 30 years of record. This criterion is more conservative 
than that used for pavements because buildings and 
other structures are less tolerant of movement than 
pavements. It is important to note that indexes found 
from weather records are for air about 4.5 feet above the 
ground; the values at ground surface, which determine 
freeze-and-thaw effects, are usually different, being 
generally smaller for freezer conditions and larger for 
thawing where surfaces are exposed to the sun. The 
surface index, which is the index determined for 
temperature immediately below the surface, is n times 
the air index, where n is the correction factor. Turf, 
moss, other vegetative cover, and snow will reduce the n 
value for temperatures at the soil surface in relation to air 
temperatures and hence give less freeze or thaw 
penetration for the same air freezing or thawing index. 
Values of n for a variety of conditions are given in TM 5- 
852-4/AFM 88-19. IChaDter~4l 

(2) More detailed information on indexes 
and th eir computa tion is presented in TM 5-852-6/AFM 
88-19, l"Chapter 6J Maps showing distribution of index 
values are presented in TM 5 -852-1 /AFM 88-19, |Chapterl 

QJ and TM 5-818-2/AFM 88-6. IChapter~4l 

c. Foundation materials. The foundation 
design decisions may be critically affected by the 
foundation soil, ice, and rock conditions. 
(1) Soils. 

(a) The most important properties of 
soils affecting the performance of engineering structures 
under seasonal freeze-thaw action are their frost-heaving 
characteristics and their shear strengths on thawing. 
Criteria for frost susceptibility based on percentage by 
weight finer than .02 millimet er are presented in TM 5- 
818-2/AFM 88-6, Chapter 4. I These criteria have also 
been developed for pavements. Heave potential at the 
lower limits of frost susceptibility determined by these 
criteria is not zero, although it is generally low to 
negligible from the point of view of pavement 
applications. Applicability of these criteria to foundation 
design will vary, depending upon the nature and 
requirements of the particular construction. Relative 
frost-heaving quali ties of vario us soils are shown in TM 
5-818-2/AFM 88-6. IChapter~4l 

(b) Permafrost soils cover the entire 
range of types from very coarse, bouldery glacial drift to 
clays and organic soils. Strength properties of frozen 
soils 



18-3 



TM 5-818-1 / AFM 88-3, Chap. 7 



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U. S. Army Corps of Engineers 

Figure 18-2. Ground temperature during freezing season in Limestone, Maine. 






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U. S. Army Corps of Engineers 

Figure 18-3. Ground temperatures during freezing season In Fairbanks, Alaska. 

18-4 



TM 5-818-1 / AFM 88-3, Chap. 7 



are dependent on such variables as gradation, density, 
degree of saturation, ice content, unfrozen moisture 
content, temperature, dissolved soils, and rate of 
loading. Frozen soils characteristically exhibit creep at 
stresses as low as 5 to 1 percent of the rupture strength 
in rapid loading. Typical strength and cr eep relation ships 
are described in TM 5-852-4/AFM 88-19. [ChapteT4l 

(2) Ice. Ice that is present in the ground in 
excess of the normal void space is most obvious as 
more or less clear lenses, veins or masses easily visible 
in cores, and test pits or excavations, but it may also be 
so uniformly distributed that it is not readily apparent to 
the unaided eye. In the annual frost zone, excess ice is 
formed by the common ice segregation process, 
although small amounts of ice may also originate from 
filling of shrinkage cracks; ice formations in this zone 
disappear each summer. Below the annual frost zone, 
excess ice in permafrost may form by the same type of 
ice segregation process as above, may occur as vertical 
ice wedges formed by a horizontal contraction-expansion 
process, or may be "fossil ice" buried by landslides or 
other events. Although most common in fine- grained 
soils, substantial bodies of excess ice are not uncommon 
in permanently frozen clean, granular deposits. The 
possible adverse effects of excess ice are discussed in 
paragraph 18-4a(2)(b). 

(3) Rock. Bedrock subject to freezing 
temperatures should never be assumed problem-free in 
absence of positive subsurface information. In seasonal 
frost areas, mud seams in bedrock or concentrations of 
fines at or near the rock surface, in combination with the 
ability of fissures in the rock to supply large quantities of 
water for ice segregation, frequently cause severe frost 
heave. In permafrost areas, very substantial quantities 
of ice are often found in bedrock, occurring in fissures 
and cracks and along bedding planes. 

d. Water conditions. 

(1) If free water drawn to developing ice 
segregation can be easily replenished from an aquifer 
layer or from a water table within a few feet of the plane 
of freezing, heave can be large. However, if a freezing 
soil has no access to free water beyond that contained in 
voids of the soil immediately at or below the plane of 
freezing, frost heave will necessarily be limited. 

(2) In permafrost areas, the supply of 
water available to feed growing ice lenses tends to be 
limited be- cause of the presence of the underlying 
impermeable permafrost layer, usually at relatively 
shallow depths, and maximum heave may thus be less 
than under otherwise similar conditions in seasonal frost 
areas. However, uplift forces on structures may be 
higher because of lower soil temperatures and 
consequent higher effective tangential adfreeze strength 
values. 

(3) The water content of soil exerts a 
substantial effect upon the depth of freeze or thaw 
penetration that will occur with a given surface freezing 
or thawing index. Higher moisture contents tend to 



reduce penetration by increasing the volumetric latent 
heat of fusion as well as the volumetric specific heat 
capacity. While an increase in moisture also increases 
thermal conductivity, the effect of latent heat of fusion 
tends to be predominant. TM 5-852-6/AFM 88-19, 
I Chapter 6l, contains charts showing thermal conductivity 
relationships. 

e. Frost-heave forces and effect of surcharge. 
Frost- heave forces on structures may be quite large. 
For some engineering construction, complete prevention 
of frost heave is unnecessary and uneconomical, but for 
most permanent structures, complete prevention is 
essential. Under favorable soil and foundation loading 
conditions, it may be possible to take advantage of the 
effect of surcharge to control heave. It has been 
demonstrated in laboratory and field experiments that the 
rate of frost heaving is decreased by an increase of 
loading on the freezing plane and that frost heaving can 
be completely restrained if sufficient pressure is applied. 
However, heave forces normal to the freezing plane may 
reach more than 10 tons per square foot. Detailed 
information on frost-heaving pressures and on the effect 
of surcharge is presented in TM 5-852-4/AFM 88-19, 

I Chapter 4J 

f. Type of structure. The type and uses of a 
structure affect the foundation design in frost areas as in 
other places. Applicabl e considera tions are discussed in 
TM 5-852-4/AFM 88-19. [ChapteT4l 

18-3. Site investigations. 

a. General. In addition to the needed site 
investigations and data described in the manuals for 
nonfrost conditions, design of foundations in areas of 
significant frost penetration requires special studies and 
data because of factors introduced by the special frost- 
related site conditions. Detailed site investigation 
procedures applicable for arctic and subarctic a reas are 
described in TM 5-8 52-2/AFM 88 -19,[ChapleL2] and TM 
5-852-4/AFM 88-19, Chapter 4, 1 and may be adapted or 
reduced in scope, as appropriate, in areas of less severe 
winter freezing. Methods of terrain evaluation in arctic 
and subarctic regions are given in TM 5-852-8. 

b. Remote sensing and geophysical 
investigations. These techniques are particularly 
valuable in selection of the specific site location, when a 
choice is possible. They can give clues to subsurface 
frozen ground conditions because of effects of ground 
freezing upon such factors as vegetation, land wastage, 
and soil and rock electrical and accoustical properties. 

c. Direct site investigations. The number and 
extent of direct site explorations should be sufficient to 
reveal in detail the occurrence and extent of frozen 



18-5 



TM 5-818-1 / AFM 88-3, Chap. 7 



strata, permafrost and excess ice including ice wedges, 
moisture contents and groundwater, temperature 
conditions in the ground, and the characteristics and 
properties of frozen materials and unfrozen soil and rock. 

(1) The need for investigation of bedrock 
requires special emphasis because of the possibilities of 
frost heave or ice inclusions as described in paragraph 
18-2c(3). Bedrock in permafrost areas should be drilled 
to obtain undisturbed frozen cores whenever ice 
inclusions could affect the foundation design or 
performance. 

(2) In areas of discontinuous permafrost, 
sites require especially careful exploration and many 
problems can be avoided by proper site selection. As an 
example, the moving of a site 50 to 100 feet from its 
planned position may place a structure entirely on or 
entirely off permafrost, in either case simplifying 
foundation design. A location partly on and partly off 
permafrost might involve an exceptionally difficult or 
costly design. 

(3) Because frozen soils may have 
compressive strengths as great as that of a lean 
concrete and because ice in the ground may be melted 
by conventional drilling methods, special techniques are 
frequently required for subsurface exploration in frozen 
materials. Core drilling using refrigerated drilling fluid or 
air to prevent melting of ice in the cores provides 
specimens that are nearly completely undisturbed and 
can be subjected to the widest range of laboratory tests. 
By this procedure, soils containing particles up to boulder 
size and bedrock can be sampled, and ice formations 
can be inspected and measured. Drive sampling is 
feasible in frozen fine-grained soils above about 250F 
and is often considerably simpler, cheaper, and faster. 
Samples obtained by this procedure are somewhat 
disturbed, but they still permit ice and moisture content 
determinations. Test pits are very useful in many 
situations. For frozen soils that do not contain very many 
cobbles and boulders, truck-mounted power augers 
using tungsten carbide cutting teeth will provide excellent 
service where classification, gradation, and rough ice- 
content information will be sufficient. In both seasonal 
frost and permafrost areas, a saturated soil condition is 
common in the upper layers of soil during the thaw 
season, so long as there is frozen, impervious soil still 
underlying. Explorations attempted during the thaw 
season are handicapped and normally require cased 
boring through the thawed layer. In permafrost areas, it 
is frequently desirable to carry out explorations during 
the colder part of the year, when the annual frost zone is 
frozen, than during the summer. 

(4) In subsurface explorations that 
encounter frozen soil, it is important that the boundaries 
of frozen and thawed zones and the amount and mode 
of ice occurrence be recorded. Materials encountered 
should be identified in accordance with the Unified Soil 



Classification Svstem TTtable 2-3 ), including the frozen soil 
classif ication syst em, as presented in TM 5-852-2/ AFM 
88-19 lChapte72l 

(5) In seasonal frost areas, the most 
essential site date beyond those needed for nonfrost 
foundation design are the design freezing index and the 
soil frost-susceptibility characteristics. In p ermafrost 
areas, as described in TM 5-852-4/AFM 88-1 9, 1 Chapter! 
[4] the date requirements are considerably more complex; 
determination of the susceptibility of the foundation 
materials to settlement on thaw and of the subsurface 
temperatures and thermal regime will usually be the 
most critical special requirements. Ground temperatures 
are measured most commonly with copper-constantan 
thermocouples or with thermistors. (6) Special site 

investigations, such as installation and testing of test 
piles, or thaw-settlement tests may be required. 
Assessment of the excavation characteristics of frozen 
materials may also be a key factor in planning and 
design. 
18-4. Foundation design. 

a. Selection of foundation type. Only sufficient 
discussion of the relationships between foundation 
conditions and design decisions is given below to 
indicate the general nature of the problems and 
sol utions. Gre ater detail is given in TM 5-852-4/AFM 88- 
19. 1 Chapter 41 

(1 ) Foundations in seasonal frost areas, 
(a) When foundation materials within 
the maximum depth of seasonal frost penetration consist 
of clean sands and gravels or other non-frost-susceptible 
materials that do not develop frost heave or thrust, or 
thaw weakening, design in seasonal frost areas may be 
the same as for nonfrost r egions, using conventional 
foundations, as indicated in Ifiqure 1 8^471 Effect of the 



frost penetration on related engineering aspects, such as 
surface and subsurface drainage systems or 
underground utilities, may need special consideration. 
Thorough investigation should be made to confirm the 
nonfrost susceptibility of subgrade soils prior to design 
for this condition. 

(b) When foundation materials within 
the annual frost zone are frost-susceptible, seasonal 
frost heave and settlement of these materials may occur. 
In order for ice segregation and frost heave to develop, 
freezing temperatures must penetrate into the ground, 
soil must be frost-susceptible, and adequate moisture 
must be available. The magnitude of seasonal heaving 
is dependent on such factors as rate and duration of 
frost penetration, soil type and effective pore size, 
surcharge, and degree of moisture availability. Frost 
heave in a freezing season may reach a foot or more in 
silts and some clays if there is an unlimited supply of 
moisture available. The frost heave may lift or tilt 



18-6 



TM 5-818-1 / AFM 88-3, Chap. 7 



SEASONAL FIOST 



PERMAFROST 



Foundations Adversely 
Affected by Freeze and Thaw 



1 i 

Foundation Wot Adversely 
Affected by Freeze or Thaw 



[Usually 
[contain: 



fine-grained soils or rock 
inlng mud seams 



3 



Use conventional foundations supported 
below annual frost zone and protected 
against uplift by adfreeze grip and 
against frost overturning or sliding 
forces 

or 
Place In the foundation compacted 
non-fro»t-susceptlble fill of 
sufficient thickness to control 
frost effects 



[Clean, granular 
soils or rock w/ 
ground ice 

Use Normal Temperate 
Zone Approach 



w/o 



Foundation Supporting Conditions 
Adversely Affected by Thaw 

[Usually fine-grained soils or] 
I rock containing ground ice Jj 



Maintenance of 

Stable Thermal Regime 

TApplicable for continuous and | 
[discontinuous permafrost zones! 



Acceptance of Thermal Regime 
Changes to be Caused by the 
Construction and Facility 

t Applicable for continuous and ~| 
discontinuous permafrost zones! 



Modification of Foundation 
Condition* Prior to Construct!. 
I" Applicable primarily for 
L discontinuous permafrost zones 



Permanent 


♦♦Temporary 


Permanent 


Temporary 


Construction 


Cooitructipn 


Construction 


Construction 


1. Piling 


1. Posts and Pads 


1. End Bearing 


1 . Piling 


2, Spread footings 


2. Sills 


piles or caissons. 


2. Perimeter 


3. Posts and Pads 


3, Slabs or rafts 


or footings, to 


■ ill* 


4. Ducted fdn. 


4, Piling 


stable stratum 




5. Refrlg. systems 




2* Rigid structural 




b. Rigid structural 




base (small 




base 




structures only) 





Permanent and Ttnmorlrr 
Conet ruction 

Use design* as applicable for 
conditions resulting after: 

1. Pre-thaw and pre-coneolidation 
of unfavorable materials, or 

2. Replacement of unfavorable 
materials 



"Permanent Construction - Construction incorporating the type and quality of materials and equipment, and details and 
methods of construction, which results in a building or facility suitable to serve a specific purpose over a minimum life 
expectancy of 25 years with normal maintenance. 

"Temporary Construction - Construction incorporating the type and quality of materials and equipment, and details and 
methods of construction, which results in a building or facility suitable to provide minimum accommodations at low first 
cost to serve a specific purpose for a short period of time, 5 years or less, in which the degree of maintenance is not a 
primary design consideration. 

Figure 18-4. Design alternatives. 



18-7 



TM 5-818-1 / AFM 88-3, Chap. 7 



foundations and structures, commonly differentially, with 
a variety of possible consequences. 

(c) When thaw occurs, the ice within 
the frostheaved soil is changed to water and escapes to 
the ground surface or into surrounding soil, allowing 
overlying materials and structures to settle. If the water 
is released by thaw more rapidly than it can be drained 
away or redistributed, substantial loss in soil strength 
occurs. In seasonal frost areas, a heaved foundation 
may or may not return to its before-heave elevation. 
Friction on lateral surface or intrusion of softened soil 
into the void space below the heaved foundation 
members may prevent full return. Successive winter 
seasons may produce progressive upward movement. 

(d) Therefore, when the soils within 
the maximum depth of seasonal frost penetration are 
frost-susceptible, foundations in seasonal frost areas 
should be supported below the annual frost zone, using 
conventional foundation elements protected against uplift 
caused by adfreeze grip and against frost overturning or 
sliding forces, or the structure should be placed on 
compacted non-frost- susceptible fill designed to control 
frost effects Kfiq 18-41 . 

(2) Foundations in permafrost areas. 
Design on permafrost areas must cope with both the 
annual frost zone phenomena described in paragraph 
18-4a(l) and those peculiar to permafrost. 

(a) Permafrost foundations not 
adversely affected by thaw. Whenever possible, 
structures in permafrost areas should be located on 
clean, non-frost-susceptible sand or gravel deposits or 
rock that are free of ground ice or of excess interstitial 
ice, which would make the foundation susceptible to 
settlement on thaw. Such sites are ideal and should be 
sought whenever possible. Foundation design under 
these conditions can be basically identical with 
temperate zone practices, even though the materials are 
frozen below the foundation support level, as has' been 
demonstrated in Corps of Engineers construction in 
interior Alaska. When conventional foundation designs 
are used for such materials, heat from the structure will 
gradually thaw the foundation to progressively greater 
depths over an indefinite period of years. In 5 years, for 
example, thaw may reach a depth of 40 feet. However, if 
the foundation materials are not susceptible to 
settlement on thaw, there will be no effects on the 
structure from such thaw. The possible effect of 
earthquakes or other dynamic forces after thawing 
should be considered. 

(b) Permafrost foundations adversely 
affected by thaw. When permafrost foundation materials 
containing excess ice are thawed, the consequences 
may include differential settlement, slope instability, 
development of water-filled surface depressions that 
serve to intensity thaw, loss of strength of frostloosened 
foundation materials under excess moisture conditions, 
development of underground uncontrolled drainage 
channels in permafrost materials susceptible to bridging 



or piping, and other detrimental effects. Often, the 
results may be catastrophic. For permafrost soils and 
rock containing excess ice, design should consider three 
alternatives, as indicated in l figure 18-4:1 maintenance of 
stable thermal regime, acceptance of thermal regime 
changes, and modification of foundation conditions prior 
to construction. Thes e approache s are discussed in TM 
5-852-4/AFM 88-19, Chapter 4. I Choice of the specific 
foundation type from among those indicated in figure 18- 
4 can be made on the basis of cost and performance 
requirements after the development of details to the 
degree needed for resolution. 

b. Foundation freeze and thaw and techniques 
for control. Detailed guidance for foundation thermal 
computations and for methods of controlling freeze-and- 
thaw penetration is presented in TM 5-852-4 and TM 5- 
852-6/AFM 88-19, Chapters 4 and 6, respectively. 

(1) Design depth of ordinary frost 
penetration. 

(a) For average permanent 
structures, the depth of frost penetration assumed for 
design, for situations not affected by heat from a 
structure, should be that which will occur in the coldest 
year in 30. For a structure of a temporary nature or 
otherwise tolerant of some foundation movement, the 
depth of frost penetration in the coldest year in 10 or 
even that in the mean winter may be used, as may be 
most applicable. The design depth should preferably be 
based on actual measurements, or on computations if 
measurements are not available. When measurements 
are available, they will almost always need to be adjusted 
by computations to the equivalent of the freezing index 
selected as the basis for design, as measurements will 
seldom be available for a winter having a severity 
equivalent to that value. 

(b) The frost penetration can be 
computed using the design freezing index and the 
detailed gui dance given in TM 5-852-6/AFM 88-19, 

I Chapter ~6~| For paved areas kept free of snow, 
approximate depths of frost pen etration m ay be 
estimated from TM 5-818-2/AF M 88-6 J ChapteT4l or TM 
852-3/AFM 88-19, iChapter 3j entering the appropriate 
chart with the air freezing index directl y. A chart i s also 
presented in TM 5-852-4/AFM 88-1 9, 1 Chapter 4| from 
which approximate depths of frost penetration may be 
obtained for a variety of surface conditions, using the air 
freezing index in combination with the appropriate 
surface index/air correction factor (n-factor). 

(c) In the more developed parts of 
the cold regions, the building codes of most cities specify 
minimum footing depths, based on many years of local 
experience; these depths are invariably less than the 
maximum observed frost penetrations. The code 



18-8 



TM 5-818-1 / AFM 88-3, Chap. 7 



values should not be assumed to represent actual frost 
penetration depths. Such local code values have been 
selected to give generally suitable results for the types of 
construction, soil moisture, density, and surface cover 
conditions, severity of freezing conditions, and building 
heating conditions that are common in the area. 
Unfortunately, the code values may be inadequate or 
inapplicable under conditions that differ from those 
assumed in formulating the code, especially for unheated 
facilities, insulated foundations, or especially cold 
winters. Building codes in the Middle and North Atlantic 
States and Canada frequently specify minimum footing 
depths that range from 3 to 5 feet. If frost penetrations 
of this order of magnitude occur with fine silt and clay- 
type soils, 30 to 100 percent greater frost penetration 
may occur in well-drained gravels under the same 
conditions. With good soil data and a knowledge of local 
conditions, computed values for ordinary frost 
penetration, unaffected by building heat, may be 
expected to be adequately reliable, even though the 
freezing index may have to be estimated from weather 
data from nearby stations. In remote areas, measured 
frost depths may be entirely unavailable. 

(2) Design depth of ordinary thaw 
penetration. Estimates of seasonal thaw penetration in 
permafrost areas should be established on the same 
statistical measurement bases as outlined in 
subparagraph a(2)(b) above for seasonal frost 
penetration. The air thawing index can be converted to a 
surface thawing index by multiplying it by the appropriate 
thawing-co nditions n-factor from TM 5-852-4/AFM 88-19, 

I Chapter 4l The thaw penetration can then be computed 
using t he detailed g uidance given in TM 5-852-6/AFM 
88-19, I Chapter ~6~1 Approximate values of thaw 
penetration may also be estimated from a chart of the air 
thawing index versus the depth of thaw in TM 5-852- 
4/AFM 88-1 9, [Chapter 4.I Degradation of permafrost will 
result if the average annual depth of thaw penetration 
exceeds the average depth of frost penetration. 

(3) Thaw or freeze beneath structures. 

(a) Any change from natural 
conditions, which results in a warming of the ground 
beneath a structure, can result in progressive lowering of 
the permafrost table over a period of years. Heat flow 
from a structure into underlying ground containing 
permafrost can only be ignored as a factor in the long- 
term structural stability when the nature of the permafrost 
is such that no settlement or other adverse effects will 
result. The source of heat may be not only the building 
heat but also the solar radiation, underground utilities, 
surface water, and groundwater flow. TM 5- 852-4 and 
TM 5-852-6/AFM 88-19, Chapters 4 and 6, respectively, 
provide guidance on procedures for estimating the depth 
of thaw under a heated building with time. 

(b) The most widely employed, 



effective and economical means of maintaining a stable 
thermal regime under a heated structure, without 
degradation of permafrost, is by use of a ventilated 
foundation. Under this scheme, provision is made for the 
circulation of cold water air between the insulated floor 
and the underlying ground. The same scheme can be 
used for the converse situation of a refrigerated facility 
supported on unfrozen ground. The simplest way of 
providing foundation ventilation is by providing an open 
space under the entire building, with the structure 
supported on footings or piling. For heavier floor 
loadings, ventilation ducts below the insulated floor may 
be used. Experience has shown that ventilated 
foundations should be so elevated, sloped, oriented, and 
configured as to minimize possibilities for accumulation 
of water, snow, ice, or soil in the ducts. Guidance in the 
thermal analysis of ventilated foundations, including the 
estimation of depths of summer thaw in supporting 
materials and design to assure winter refreezing, is given 
TM 5-852-4 and TM 5-852-6/AFM 88-19, Chapters 4 and 
6, respectively. 

(c) Natural or forced circulation 
thermal piles or refrigeration points may also be used for 
overall foundation cooling and control of permafrost 
degradation. 

(4) Foundation insulation. Thermal 
insulation may be used in foundation construction in both 
seasonal frost and permafrost areas to control frost 
penetration, frost heave, and condensation, to conserve 
energy, to provide comfort, and to enhance the 
effectiveness of foundation ventilation. Unanticipated 
loss of effectiveness by moisture absorption must be 
avoided. Cellular glass should not be used where it will 
be subject to cyclic freezing and thawing in the presence 
of moisture. Insulation thicknesses and placement may 
be determined by the guidance given in TM 5-852-4 and 
TM 5-852-6/AFM 88-19, Chapters 4 and 6, respectively. 

(5) Granular mats. In areas of significant 
seasonal frost and permafrost, a mat of non-frost- 
susceptible granular material may be used to moderate 
and control seasonal freeze-and-thaw effects in the 
foundation, to provide drainage under floor slabs, to 
provide stable foundation support, and to provide a dry, 
stable working platform for construction equipment and 
personnel. Seasonal freezing-and-thawing effects may 
be totally or partially contained within the mat. When 
seasonal effects are only partially contained, the 
magnitude of seasonal frost heave is reduced through 
both the surcharge effect of the mat and the reduction of 
frost penetration into u nderlying fro st-susceptible soils. 
TM 5-852-4/AFM 88-19, Chapter 4. I provides guidance in 
the design of mats. 

(6) Solar radiation thermal effects. The 
control of summer heat input from solar radiation is very 
important in foundation design in permafrost areas. 
Correc- 



18-9 



TM 5-818-1 / AFM 88-3, Chap. 7 



tive measures that may be employed include shading, 
reflective paint or other surface material, and sometimes 
live vegetative covering. In seasonal frost areas, it may 
sometimes be advantageous to color critical surfaces 
black to gain maximum effect of solar heat in reducing 
winter frost problems. TM 5-852-4/AFM 88-19, lChapterl 
l"4] provides guidance on the control of solar radiation 
thermal effects. 

c. Control of movement and distortion. The 
amount of movement and distortion that may be 
tolerated in the support structure must be established 
and the foundation must be designed to meet these 
criteria. Movement and distortion of the foundation may 
arise from seasonal upward, downward, and lateral 
displacements, from progressive settlement arising from 
degradation of permafrost or creep deflections under 
load, from horizontal seasonal shrinkage and expansion 
caused by temperature changes, and from creep, flow, 
or slide of material on slopes. Heave may also occur on 
a nonseasonal basis if there is progressive freezing in 
the foundation, as under a refrigerated building or 
storage tank. If the subsurface conditions, moisture 
availability, frost penetration, imposed loading, or other 
factors vary in the foundation area, the movements will 
be nonuniform. Effects on the foundation and structure 
may include various kinds of structural damage, jamming 
of doors and windows, shearing of utilities, and problems 
with installed equipment. 

(1) Frost-heave and thaw-settlement 
deformations. 

(a) Frost heave acts in the same 
direction as the heat flow, or perpendicular to the 
freezing plane. Thus, a slab on a horizontal surface will 
be lifted directly upward, but a vertical retaining wall may 
experience horizontal thrust. Foundation members, such 
as footings, walls, piles, and anchors, may also be 
gripped on their lateral surface s and heaved by frost 
forces acting in tangential shear. I Figure IS^BI shows an 
example of frost-heave forces developed in tangential 
shear on timber and steel pipe piles restrained against 
upward movement. 

(b) In rivers, lakes, or coastal water bodies, 
foundation members to which floating ice may adhere 
may also be subject to important vertical forces as water 
levels fluctuate. 

(c) Among methods that can be used to control 
detrimental frost action effects are placing non- 
frostsusceptible soils in the depth subject to freezing to 
avoid frost heave or thrust; providing sufficient 
embedment or other anchorage to resist movement 
under the lifting forces; providing sufficient loading on the 
foundation to counterbalance upward forces; isolating 
foundation members from heave forces; battering or 
tapering members within the annual frost zone to reduce 
effectiveness of heave grip; modifying soil frost 
susceptibility; in seasonal frost areas only, taking 
advantage of natural heat losses from the facility to 



minimize adfreeze and frost heave; or cantilevering 
building attachments, e.g., porches and stairs, to its main 
foundation. 

(d) In permafrost areas, movement 
and distortion caused by thaw of permafrost can be 
extreme and should be avoided by designing for full and 
positive thermal stability whenever the foundation would 
be adversely affected by thaw. If damaging thaw 
settlement should start, a mechanical refrigeration 
system may have to be installed in the foundation or a 
program of continual jacking may have to be adopted for 
leveling of the structure. Discontinuance or reduction of 
building heat can also be effecti ve. Detaile d guidance is 
given in TM 5-852-4/AFM 88-19. IUFiapteT4l 

(2) Creep deformation. Only very small 
loads can be carried on the unconfined surface of ice- 
saturated frozen soil without progressive deformation. 
The allowable long-term loading increases greatly with 
depth but may be limited by unacceptable creep 
deformation well short of the allowable stress level 
determined from conventional short-term test. Present 
practice is to use large footings with low unit loadings; 
support footings on mats of well-drained non-frost- 
susceptible granular materials, which reduce stresses on 
underlying frozen materials to conservatively low values; 
or place foundations at sufficient depth in the ground so 
that creep is effectively minimized. Pile foundations are 
designed to not exceed sustainable adfreeze bond 
strengths. In all cases, analysis is based on permafrost 
temperature at the warmest time of the year. For cases 
which require estimation o f foundation creep behavior, 
see TM 5-852-4/AFM 88-1 9, 1 Chapter 7 ?] 

d. Vibration problems and seismic effects. 

(1) Foundations supported on frozen 
ground may be affected by high stress-type dynamic 
loadings, such as shock loadings from high-yield 
explosions, by lower stress pulse-type loadings as from 
earthquakes or impacts, or by relatively low-stress, 
relatively low-frequency, steady-state vibrations. In 
general, the same procedures used for nonfrozen soil 
conditions are applicable to frozen s oils. Design criteria 
are given in TM 5-809-1 0/AFM 88-3, IChapter 1B ; TM 5- 
856-4; TM 5-852-4/AFM 88-19, Chapter 4:l and Ichac-terl 
QJ' of this manual. These manuals also contain 
references to sources of data on the general behavior 
and properties of nonfrozen soils under dynamic load 
and discuss types of laboratory and field tests available. 
However, design criteria, test techniques, and methods 
of analysis are not yet firmly established for engineering 
problems of dynamic loading of foundations. Therefore, 
the Office, Chief of Engineers, ATTN: DAEN- ECE-T, 
WASH DC 20314, or HQUSAF/PREE, WASH DC 
20332, should be notified upon initiation of design and 
should participate in establishing criteria and ap- 



18-10 



TM 5-818-1 / AFM 88-3, Chap. 7 



20 



in 
a. 



—1— 


1 ' 1 


1 | ! | . | 1 | I | 1 | 1 | I | 1 | 1 | 1 


1 1 1 1 1 1 1 1 1 


1 ' 1 


1 l 


— 




1958-1959 

Seasonal thaw layer extended below bottom of pile, 
affecting heave force measurements. 


y 19 62 -1963 




- 


- 


/ 








\- 



10 — 




8 STEEL PIPE PILES 



U.S. Army Corps of Engineers 



AVERAGE ADFREEZE STRESS vs TIME 



Figure 18-5. Heave force tests, average tangential adfreeze bond stress versus time, and timber and steel pipe piles 
placed with silt-water slurry in dry excavated holes. Piles were installed within annual frost zone only, over permafrost, to 

depths from ground surface of 3.6 to 6.5 feet. 



18-11 



TM 5-818-1 / AFM 88-3, Chap. 7 



proach and in planning field and laboratory tests. 

(2) All design approaches require knowledge of 
the response characteristics of the foundation materials, 
frozen or nonfrozen, under the particular load involved. 
As dynamic loadings occur in a range of stresses, 
frequencies, and types (shock, pulse, steadystate 
vibrations, etc.), and the response of the soil varies 
depending upon the load characteristics, the required 
data must be obtained from tests that produce the same 
responses as the actual load. Different design criteria 
are used for the different types of dynamic loading, and 
different parameters are required. Such properties as 
moduli, damping ability, and velocity of propagation vary 
significantly with such factors as dynamic stress, strain, 



frequency, temperature, and soil type a nd condition. TM 
5-852-4/AFM 88-19, Chapter 4~l discusses these 
properties for frozen ground. 

e. Design criteria for various specific 
engineering features. In addition to the basic 
considerations outlined in the preceding paragraphs of 
this chapter, the design of foundations for frost and 
permafrost conditions requires application of detailed 
criteria for specific engineering situations. Guidance for 
the design of various specific features, construction 
consideration, and monitoring of performance of 
foundation is presented in TM 5-852-4/AFM 88-19, 
I Chapter 4J 



18-12 



TM 5-818-1 / AFM 88-3, Chap. 7 



APPENDIX A 
REFERENCES 



Government Publications 



Departments of the Army, the Navy, and the Air Force 



TM 5-809-7 Pile Foundations 
TM 5-809-1 0/AFM 88-3, Chap. 13 
TM 5-818-2/AFM 88-6, Chap. 4 
TM 5-818-4/AFM 88-5, Chap. 5 
TM 5-818-5/AFM 88-5, Chap. 6 
TM 5-81 8-6/AFM 88-32 
TM 5-818-7 

TM 5-824-3/AFM 88-6, Chap. 3 
TM 5-852-1 /AFM 88-19, Chap. 1 
TM 5-852-2/AFM 88-19, Chap. 2 

TM 5-852-3/AFM 88-19, Chap. 3 
TM 5-852-4/AFM 88-19, Chap. 4 
TM 5-852-5/AFM 88-19, Chap. 5 
TM 5-852-6/AFM 88-19, Chap. 6 

TM 5-852-7/AFM 88-19, Chap. 7 

TM 5-852-8 

TM 5-852-9/AFM 88-19, Chap. 9 
TM 5-856-4 

NAVFAC DM-7 



Seismic Design for Buildings 

Soils and Geology: Pavement Design for Frost Conditions 

Soils and Geology: Backfill for Subsurface Structures 

Dewatering and Groundwater Control for Deep Excavations 

Grouting Methods and Equipment 

Foundations on Expansive Soils 

Rigid Pavements for Airfields Other Than Army 

Arctic and Subarctic Construction: General Provisions 

Arctic and Subarctic Construction: Site Selection and 

Development 
Arctic and Subarctic Construction: Runway and Road Design 
Arctic and Subarctic Construction: Building Foundations 
Arctic and Subarctic Construction: Utilities 
Arctic and Subarctic Construction: Calculation Methods for 

Determination of Depths of Freeze and Thaw in Soils 
Arctic and Subarctic Construction: Surface Drainage Design 

for Airfields and Heliports in Arctic and Subarctic Regions 
Arctic and Subarctic Construction: Terrain Evaluation in 

Arctic and Subarctic Regions 
Arctic and Subarctic Construction: Buildings 
Design of Structures to Resist the Effects of Atomic Weapons: 

Structural Elements Subjected to Dynamic Loads 
Soil Mechanics, Foundations, and Earth Structures 



5-818-1 / AFM 88-3, Chap. 7 



The proponent agency of this publication is the Office of the Chief of Engineers, United States 
Army. Users are invited to send comments and suggested Improvements on DA Form 2028 
(Recommended Changes to Publications and Blank Forms) direct to HQDA (DAEN-ECE-G), WASH 
DC 20314. 



By Order of the Secretaries of the Army and the Air Force: 



Official: Chief of Staff 

ROBERT M. JOYCE 
Major General, United States Army 
The Adjutant General 



JOHN A. WICKHAM, JR. 
General, United States Army 



CHARLES A. GABRIEL, General, USAF 



Official: 



Chief of Staff 



JAMES H. DELANEY, Colonel, USAF 
Director of Administration 



Distribution: 

Army: To be distributed in accordance with DA Form 12-34B, requirements for TM 5-800 Series: Engineering 
and Design for Real Property Facilities. 

AirForce: F 



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